The Foundation Engineering Handbook Chapter 6

The Foundation Engineering Handbook Chapter 6 Geotechnical earthquake engineering can be defined as that subspecialty within the field of geotechnical engineering that deals with the design and construction of projects in order to resist the effects of earthquakes. Geotechnical earthquake engineering requires an understanding of basic geotechnical principles as well as an understanding of geology, seismology, and earthquake engineering. In a broad sense, seismology can be defined as the study of earthquakes. This would include the internal behavior of the earth and the nature of seismic waves generated by the earthquake.

6.2.1 Selection of Pile Material for Different Construction Situations 236

6.3 Estimation of Static Pile Capacity of a Single Pile 239

6.3.3.1 Pile Capacity Estimation from Standard Penetration Test Results 2476.3.3.2 Pile Capacity Estimation from Cone Penetration Test Results 248

6.6.1 Elastic Solution 2596.6.2 Computation of Pile Settlement Using Approximate Methods 262

Page 236

6.7.3 Approximate Methods for Computation of Immediate Settlement of Pile

Groups

271

in piling, this foundation type is only utilized after other less costly alternatives, such as (1)combined footings and (2) ground modifications, have been considered and ruled out for theparticular application On the other hand, piles may be the only possible foundation

6.2 Design of Pile Foundations6.2.1 Selection of Pile Material for Different Construction Situations

Depending on applicability in a given construction situation, one of three different pile types,timber, concrete, or steel, is selected to construct a pile foundation

6.2.1.2 Concrete Piles

Concrete piles can be selected for foundation construction under the following circumstances:

1 The need to support heavy loads in maritime areas where steel piles easily corrode

2 Existence of stronger soil types located at relatively shallow depths that are accessible toconcrete piles

3 Design of bridge piers and caissons that require large-diameter piles

4 Design of large pile groups is needed to support heavy extensive structures so that the totalexpense can be minimized

5 The need for minipiles to support residential buildings on weak and compressible soils

FIGURE 6.1

(a) Groups of timber piles in construction (From www.timberpilingcouncil.org With permission.) (b)

Production of precast concrete piles (From www.composite-piles-marine-pilings.com With permission.) (c) Steel sheet piles in a cofferdam application (From www.dissen- juhn.com With permission.)

The disadvantages of concrete piles are that they can be damaged by acidic environments ororganic soils and they undergo abrasion due to wave action when used to construct offshorefoundations.

Concrete piles are in wide use in construction due to their relatively high capacity and

reasonable cost The two most common types of concrete piles are (1) precast and (2)

castin-situ Of these, precast piles may be constructed to specifications at a separate casting yard or

at the pile construction site itself if a large number of piles are needed for the particular

construction In any case, handling and transportation can cause intolerable tensile stresses inprecast concrete piles Hence, one should be cautious in handling and transportation so as tominimize the bending moments in the pile Two other important issues that have to be

addressed with precast piles that have to be driven are the ground displacement that theycause and the possible damage due to driving stresses Therefore, driving of precast pileswould not be suitable for construction situations where soildisplacement-sensitive structuresare located in the proximity Preaugering or jetting would be alternative installation

techniques to suit such construction situations

Cast-in-situ piles are of two types:

1 Cased type, which are piles that are cast inside a steel casing that is driven into the ground

2 Uncased type, which are piles that are formed by pouring concrete into a drilled hole or into

a driven casing before the casing is gradually withdrawn

A detailed discussion of the use of casings in cast-in-situ pile construction is found in Bowles

(2002)

Auger-cast concrete piles have the following properties:

• Higher capacity having larger diameter (tall building foundation)

• Low vibration during construction (business districts with high-rise buildings)

• Higher depth (load transfer into deeper strong soil)

• Replacement pile (no lateral soil movement) No compression of surrounding soil

6.2.1.3 Steel Piles

Steel piles offer excessive strength in both compression and tension In addition, they arehighly resistant to structural damage during driving Furthermore, they can be spliced veryconveniently to suit any desired length On the other hand, the main disadvantages of steelpiles are (1) high expense and (2) vulnerability to corrosion in marine environments

Therefore, steel piles are ideal for supporting excessively heavy structures such as multistoreybuildings in soft ground underlain by dense sands, stiff clays, or bedrock in nonmarine

environments

6.2.2 Selection of the Method of Installation

Piles can be classified into three categories depending on the degree of soil displacementduring installation: (i) large volume displacement piles; (ii) small volume displacement piles;and (iii) replacement piles Driven precast solid concrete piles, close ended pipe piles, anddriven and cast in-place concrete piles fall into the large volume displacement category inwhich a large volume of soil is displaced during installation Steel piles

Page 239

with thin cross sections, for example, H and open-ended pipe piles, fall into the small volumedisplacement pile category where the amount of soil displaced during installation is small Allbored and cast in-place concrete piles and caissons fall into the replacement pile category, inwhich the soil is removed and replaced with concrete Installation of large volume

displacement piles obviously causes disturbance to the soil surrounding the pile

6.2.3 Design Criteria

Failure of a structurally intact pile can be caused due to two reasons: (1) shear failure of thesoil surrounding the pile and (2) excessive settlement of the foundation Therefore, the task ofthe foundation designer is to find out an economical pile to carry the working load with a lowprobability of shear failure, while keeping the resulting settlement to within allowable limits

In designing a single pile against shear failure, it is customary to estimate the maximum loadthat can be applied to a pile without causing shear failure, generally referred to as the ultimatecarrying capacity

As in the case of shallow footings, two design approaches, (1) allowable stress design(ASD) method and (2) load resistance factor design (LRFD) method, are available for piles.The following sections will mostly elaborate the ASD method and basics of the LFRD

method will be presented in Section 6.9 The ASD requires the following conditions:

6.3 Estimation of Static Pile Capacity of a Single Pile

Piles are usually placed in service as a group rather than on an individual basis to meet

loading demands and ensure stability In addition, if some probability of nonvertical loadingalso exists and the designer is uncertain of the lateral capacity of the piles, then it is common

to include some battered piles as well in the group (Figure 6.2)

As one realizes from Figure 6.2, the structural load (Pstructural) is transferred to each

individual pile in the group (Ppile,i) through the pile cap The relation between Pstructuraland

Ppile,iis determined by considering the pile cap as a statically determinate or a statically

individual pile capacity, can meet the demand of the load imposed on it, i.e., Ppile,i.

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FIGURE 6.2

Piles in a typical service condition.

The pile designer must be knowledgeable of the capacity of a pile (1) under normal workingconditions (static capacity) and (2) while it is driven (dynamic capacity) Since the dynamicpile capacity is addressed in detail inChapter 8, discussions in this chapter would be limited

to the static pile capacity only (Figure 6.3)

The ultimate working load that can be applied to a given pile depends on the resistance thatthe pile can produce in terms of side friction and point bearing (Figure 6.2) Hence, the

expression for the allowable load Paon a pile would take the following form:

(6.3)

where Ppuis the ultimate point capacity, Psuis the ultimate side friction, and FS is the safetyfactor

A suitable factor of safety is applied to the ultimate carrying capacity to obtain the

allowable load on a pile, subject to the allowable settlement The magnitude of the safety

FIGURE 6.3

Illustration of pile capacity.

factor depends on the confidence of the designer on the design, and a factor of safety between

3 and 4 is very often used

6.3.1 Estimation of Point Capacity

6.3.1.1 Meyerhoff’s Method

The ultimate point capacity component in Equation (6.3) corresponds to the bearing capacity

of a shallow footing expressed by Equation (3.1), and is a modified form of Equation (3.2):

(6.4)

where A p is the area of the pile cross section, q is the vertical effective stress at the pile tip, c

is the cohesion of the bearing layer, and and are the bearing capacity factors modified for

deep foundations (and a B/L ratio of 1.0).

It is noted that the surcharge component (0.5BNγ.γ) of Equat ion (3.2) ha s been due to theinsignificance of the surcharge zone of the pile compared to the entire stress regime along thedepth of the pile

The bearing capacity factors for deep foundations can be found inFigure 6.4 However, use

of the bearing capacity factors mentioned above is more complex than in the case of shallowfootings since, in the case of deep foundations, the mobilization of shear strength also

depends on the extent of the pile’s penetration into the bearing layer In granular soils, thedepth ratio at which the maximum strength is mobilized is called the critical depth ratio

(Lb/D)crfor the mobilization and for different values of Φ(Figure 6.5)

According to Meyerhoff (1976), the maximum values of and are usually mobilized at

depth ratios of 0.5 (Lb/D)cr Hence, one has to follow an interpolation process to evaluate the

bearing capacity factors if the depth ratio is less than 0.5 (Lb/D)cr This is illustrated in

Example 6.1

FIGURE 6.4

Bearing capacity factors for deep foundations.

where cuis the undrained strength of the clay It must be noted that in the case of steel piles

(HP or pipe type) Apis usually computed as the entire cross-sectional area due to plugging ofthe section with bearing soil, especially when it is driven to firm bearing However, when

piles are driven to bearing on rock, Apis computed as the steel area of the cross section

6.3.1.2 Vesic’s Method

Based on the theory of cavity expansion, Vesic (1977) expressed the point bearing capacity of

a pile by an expression similar to Equation (6.4):

where

c and q are defined as in Equation (6.4) K0is the coefficient of earth pressure at rest and

be obtained fromTable 6.1(a)and (b) based on the rigidity index Irand the reduced rigidity

index Irrdefined as follows:

recommended for Ir(Bowles, 2002):

It is noted that in the case of sand that does not exhibit volumetric dilation or undrainedclay

6.3.2 Skin-Friction Capacity of Piles

The skin-friction capacity of piles can be evaluated by means of the following expression:

(6.11)

where p is the perimeter of the pile section, z is the coordinate axis along the depth direction, f

is the unit skin friction at any depth z, and L is the length of the pile.

6.3.2.1 Unit Skin Friction in Sandy Soils

Since the origin of skin friction in granular soils is due to the frictional interaction betweenpiles and granular material, the unit skin friction (skin-frictional force per unit area) can beexpressed as

(6.12)

where K is the earth pressure coefficient (K0for bored piles and 1.4 K0for driven piles), δis

the angle of friction between the soil and the pile material (usually assumed to be 2/3 if onelooks for a generic value; if a more appropriate value for interaction between a particular pilematerial and a soil is needed, one can use the values suggested inChapter 10), and is the

vertical effective stress at the point of interest (i.e., where f is compute

It can be seen from the above expression that the unit skin friction can increase linearlywith depth However, practically, a depth of 15B (where B is the cross-sectional dimension)

has been found to be the limiting depth for this increase K0, the coefficient of lateral earth

pressure at rest, is typically expressed by

(6.13)

6.3.2.2 Skin Friction in Clayey Soils

In clayey soils, on the other hand, skin friction results from adhesion between soil particlesand the pile Hence, the unit skin friction can be simply expressed by

Undrained Strength (kPa) a

Computation of skin friction in loose sand

Applying Equations (6.12) and (6.13), one would obtain

f=1.4K0(17.0)z tanδ

up to a depth of −3.0m (i.e., 15.0×0.2) and constant thereafter

Assume that

Also assume that the surficial loose sand is normally consolidated Hence the over

consolidation ratio (OCR)=1.0, K0=0.574

f=4.203z kPa for z < 3m

f=12.6 kPa for z > 3m

Computation of skin friction in clay

Applying Equation (6.14), one obtains

Psf=(0.8)[0.5(3)(12.6)+12.6(1)+40(6)]=217.2 kN

Computation of the point resistance in dense sand

FromFigure 6.5, (L/D)cr=15 For the current problem, L/D=1/0.2=5 Since in this

case,

FIGURE 6.6

Illustration for the Example 6.1

Note that an value of 300 is obtained fromFigure 6.4 Also

Then, by substituting in Equation (6.5)

In practice, the ultimate carrying capacity is estimated using the static bearing capacity

methods and then often verified by pile load tests

6.3.2.3 The αMethod

In the author’s opinion, the pile capacity evaluations outlined above are generalized in the

method popularly known as the a method expressed as follows:

(6.15a)

in which the mathematical symbols have been defined based on Equations (6.12) and (6.14).Sladen (1992) derived the following analytical expression that explains the dependence ofthe a factor on the undrained shear strength of saturated fine grained soils

(6.15b)

where suis the undrained shear strength described in the Section 1.4.2.2and C1=0.4 to 0.5 forbored piles and greater than 0.5 for driven piles

6.3.2.4 The βMethod

The βmethod suggested by Burland (1973) for the computation of skin-friction derives from

the concepts used in the formulation of Equation (6.12) that is used for the determination ofskin friction in granular soils It can be expressed in the following general formulation:

Comparison of Equations (6.12) and (6.16) shows that the factor βrepresents the term K tan θ ,

which is completely dependent on the angle of friction Bowles (2002) shows that,

for most granular soils, the factor βis in the range of 0.27 to 0.3, providing a convenient

practical way of evaluating the skin friction of piles in granular soils

6.3.2.5 The λ Method

A semiempirical approach for prediction of skin-friction capacity of piles in clayey soils waspresented by Vijayvergia and Frocht (1972) based on load tests conducted on long piles thatsupport offshore oil production structures The corresponding expression for skin-frictioncapacity is given in Equation (6.17)

(6.17)

Based on back calculation of observed capacities of static pile load tests, the nondimensionalcoefficient λ has been presented as a function of the depth as shown in Figure 6.7

6.3.3 Pile Capacity Estimation from In Situ Tests

6.3.3.1 Pile Capacity Estimation from Standard Penetration Test Results

Meyerhoff (1976) proposed a relationship (Equation (6.18)) to determine the point capacity of

a pile in coarse sand and gravel, in kPa, using standard penetration test (SPT) data:

As pointed out inSection 6.3.1, the point resistance reaches a limiting value at a critical

Lb/D For the above outlined qpuvs N relationship (Equation (6.18)), the suggested critical

Lb/D is about 90 Meyerhoff (1976) also proposed the following alternative relationship for

nonplastic silt:

qpu=300N

(6.19)

FIGURE 6.7

Dependence of the λ factor on pile penetration.

On the other hand, for the ultimate unit skin friction in sands, the following relationships wereproposed by Meyerhoff (1976):

6.3.3.2 Pile Capacity Estimation from Cone Penetration Test Results

AASHTO (1996) recommends the following technique proposed by Nottingham and

Schmertmann (1975) to determine the point bearing capacity in clay based on cone

penetration data:

(6.21)

where qc1 and qc2are minimum averages (excluding sudden peaks and troughs) of qc values inthe influence zones below the pile tip and above the pile tip, respectively These influencezones are shown in Figure 6.8 R1is a reduction factor evaluated fromTable 6.3 R2is 1.0 forthe electrical cone and 0.5 for the mechanical cone

A similar expression is available for the evaluation of point bearing resistance of sands(DeRuiter and Beringen, 1979):

(6.22)

where qc1 and qc2are minimum averages (excluding sudden peaks and troughs) of qc values inthe influence zones below the pile tip and above the pile tip, respectively for normallyconsolidated sand and 0.67 for overconsolidated sand

FIGURE 6.8

Tip influence zone.

fsu=α 'fs

(6.23)

In the case of electrical cone penetrometers, a', the frictional resistance modification factor

can be evaluated from Table 6.4based on the depth of embedment, Z/B

Tomlinson (1994) advocates the use of the cone resistance in evaluating the skin frictiondeveloped in piles since the former is found to be more sensitive to variations in soil densitythan the latter Tomlinson (1994) provides the empirical data inTable 6.5for this evaluation

Example 6.2

The SPT profile of a site is shown inFigure 6.9 Estimate the depth to which a HP 360×

108 pile must be driven at this site if it is to carry a load of 1500 kN Assume that the SPT testwas performed in silty clay in the absence of water and the unit weights of peat, silty clay(dry), saturated silty clay, and saturated medium-dense sand are 10.5, 16.0, 17 5, 17.2 kN/m3,

respectively Use Meyerhoffs method for estimating point bearing and the αmethod for

estimating skin-friction capacity

TABLE 6.4

Frictional Resistance Modification Factors Applied to CPT Results (α ' )

35 0.8 0.7 0.75

TABLE 6.5

Relationships between Pile Shaft Friction and Cone Resistance

Open-ended steel tube driven into fine to medium sand 0.0033qc

Source: From Tomlinson, M.J., 1994, Pile Design and Construction Practices, 4th ed., E & FN Spon, London.

Minimum pile dimension=0.346 m

Limiting skin-friction depth in sand=Ls,lm=15(0.346)=5.19m (assumed to apply

from the clay-sand interface)

Critical end-bearing penetration=Lp,cr(Figure 6.5)=3(0.346) for clays=1 m

=10(0.346) for sand=3.46m

The following soil strength properties can be obtained based on the SPT values (Table 6.6):

value are obtained fromFigure 6.4

δ=2/3 N

K=1.4K0tan δ=1.4(1−sin N) tan δ

From Equations (6.7) and (6.14), the maximum total ultimate resistance produced by theclayey layers=point bearing+skin friction=9(0.128)(25)+(2.2)[1.03(19)(1) +

0.92(50)(4)+(1.0)(25)(2)]=28.8+556.6–585.4 kN

Hence, the pile has to be driven into sand (say up to a depth of L m).

TABLE 6.6

Soil Parameters Related to the Pile Design in Example 6.2

7–10.1 34 100 23 0.26 No 3.46

FIGURE 6.9

Illustration for Example 6.2

Since the critical embedment is 3.46 m, one can assume that the pile needs to be driven

passing a 12.3 m depth for complete mobilization of point capacity and skin friction

Assume that for depths greater than 16.8 m the soil properties are similar to those from 12.3

to 16.8m

Effective clay overburden=(10.5)(1)+16(4)+(17.5–9.8)(2)=89.9 kPa

Effective sand overburden=(L−7)(17.2–9.8)=7.4L−51.8

From Equation (6.5) for net ultimate point resistance

A cased concrete pile is required to carry a safe working load of 900 kN in compression at

a site where the CPT results are given inFigure 6.10 Recommend a suitable pile size and adepth of penetration

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FIGURE 6.10

CPT results for Example 6.3

FromFigure 6.10(a), it is seen clearly that the immediate subsurface consists of a loose sandlayer up to a depth of 11.0 m underlain by a denser sand layer

Based on the cone resistance (Figure 6.10), and Equation (6.22), the maximum endbearingresistance that can be obtained from the loose fine sand layer is

Assume that a 400 mm diameter pile is employed in order not to overstress the concrete asshown later in the solution The tip area of this pile=0.126m2and the pile perimeter= 1.26m.Then, the maximum working load that can be carried at the tip is computed as

Computational Aid for Example 6.4

Depth Interval (m) fs (MPa) αfs (kPa) Psu (kN), Equation (6.11)

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For a depth of embedment (z/B) ratio of 13.5/0.4=33.75, fromTable 6.4, α=0.7 As shown inTable 6.7, Equation (6.23) can be applied on incremental basis,

Total ultimate skin friction=1654 kN

Static pile capacity=(1260+1654)/2.5=1165.6kN

Hence, the load can be carried safely at a pile embedment of 13.0 m The same design isrepeated under LRFD guidelines in Section 6.9

6.4 Pile Load Transfer

When a structural load is applied on a pile, it will be supported by certain amounts of skinfriction and point bearing resistance that are mobilized as required The degree of

mobilization of both skin friction and point bearing resistance depends on the relative

displacement undergone by the pile at the particular location of reference with respect to thesurrounding soil This condition is graphically illustrated in Figure 6.11, that shows the

mobilization of skin friction through shear stress along points on the embedded pile surfacegoverned by the shear strain undergone by the pile with respect to the surrounding soil (slip)

at those locations Although the magnitude of slip needed to mobilize the ultimate shearresistance depends on the soil type, typically it would be within a few millimeters (e.g.,

<10mm) Similarly, mobilization of point (tip) resistance depends on the axial strain or

penetration of the pile tip in the bearing layer and for complete mobilization, a penetration of

10 to 25% of the pile diameter would be required

The discussion in the previous section enables one to evaluate the ultimate or the maximumresistance that can be mobilized at the tip or the shaft It is quite typical of many soil types toreach a critical state at a much higher strain than is required for the mobilization of ultimate orpeak strength, especially in shear The shear strength at the critical state is known as theresidual shear strength (Figure 6.11)

Based on the above discussion, one realizes that when a certain structural load is applied on

a pile that has been already installed by driving or in situ casting, the following conditions

must be satisfied:

1 Pworking= Ps+Pp

2 Psand Ppcause an immediate settlement of the pile with respect to the surrounding layersand the bearing layer producing slip at the frictional interface and penetration of the bearing

layer, respectively The above slip induces shear strains, γ , and the penetration induces an

axial strain, ε , on the pile tip.

3 The magnitudes of γand ε , respectively, determine the levels of interfacial shear stress,

normal stress at the tip, σ, based on the deformation characteristics

FIGURE 6.11

Page 254

shown inFigure 6.11 On the other hand, it is the mobilization of and σthat finally

determine the magnitudes of Psand Pp

It is realized how an interplay between forces Psand Ppoccurs under conditions 1 to 3 until

an equilibrium is finally reached This process is known as the pile load transfer process Atypical load transfer curve at the equilibrium is illustrated inFigure 6.12 A load transfercurve such as inFigure 6.12 depicts the axial load carried by the pile at any given depth.Hence, the difference between the applied load and the axial load at that depth indicates thecumulative frictional resistance mobilized up to that depth The axial load in the pile effective

at any depth z can be experimentally determined by installing strain gages at that depth When

the longitudinal strain, εz at a depth z, is electronically monitored, the axial force at that point,

P(z), can be estimated as follows:

P(z)=EApεz

(6.24)

where E is the elastic modulus of the pile material.

The plot of P(z) vs z (the load transfer curve) inFigure 6.12corresponds to the applied

working load of Pw1where the mobilized point resistance is shown as Pp1 The above

technique also provides one with the means of observing the variation of the load transfer

curve as Pwis increased, for instance, to Pw2(Figure 6.12)

6.5 Time Variation of Pile Capacity (Pile Setup)

Due to the initial disturbance caused by pile installation and the consequent stabilization ofthe surrounding soil and pore water; in most soils, the axial or lateral pile capacity changeswith time A number of researchers have proposed analytical methods to estimate the change

in pile capacity with time Thilakasiri et al (2003) have performed a case study comparingmany previously established methods of predicting variation of pile capacity with time

FIGURE 6.12

Pile load transfer curves.

Flaate (1972) identified three different zones surrounding a driven pile in clay: (i) remoldedzone of 100 to 150 mm thickness from the pile surface; (ii) transition zone; and (iii)

unaffected zone outside the transition zone Pile driving can set up high pore pressure in theremolded zone and the soil is remolded under constant water content The pore pressuregenerated during the installation process would be dissipated with time depending on thepermeability of the surrounding soils Further, the structure of the soil disturbed due to drivingmay also be recovered with time The process of recovery of the soil structure with time andthe consolidation of surrounding soil with time due to dissipation of the excess pore pressure

is termed “thixotropic recovery” It is believed that the time taken for the recovery to be

complete depends on the amount of disturbance caused by the pile installation process and theproperties of the surrounding soil

Due to the thixotropic recovery, the ultimate carrying capacity of the pile will vary withtime If the ultimate carrying capacity of the pile is increased due to thixotropic recovery, it istermed “set up”, whereas if it is decreased, it is termed “relaxation.” The phenomenon oftime-dependent strength gain in piles driven into cohesive soil deposits is well established(Fellenius et al., 1989; Skove et al, 1989; Svinkin and Skov, 2002)

Due to the rapid dissipation of excess pore pressure, the increase in bearing capacity ofpiles driven into sandy deposits is expected to be complete within a few hours or at mostwithin a few days after installation However, substantial increases in capacity of driven piles

in sand over a long period of time have been reported (Tavenas and Audy, 1972; York et al.,1994; Tomlinson, 1994; Chun et al., 1999) Since a substantial increase in the ultimate

carrying capacity of a pile driven into sand over a long period of time cannot be attributed todissipation of excess pore pressure, Chun et al (1999) suggested the possibility of otherreasons for such an increase in the ultimate carrying capacity Some of these are: (i) bonding

of sand particles to the pile surface; (ii) increase in strength due to soil aging; and (iii) term changes in the stress state surrounding the pile due to breakdown of arching around thepile resulting from the creep behavior of sand particles

long-Svinkin and Skov (2002) modified an earlier relationship suggested by Skov and Denver(1989) for cohesive soils to include the pile capacity at the end of initial driving (EOID)

(6.25a)

where Ru(t) is the bearing capacity of the pile at time t, t is the time since end of initial driving

(EOID), REOIDis the bearing capacity of the pile at the end of initial driving, B is a factor,

depending on soil type, pile type, and size evaluated by fitting field data

Svinkin and Skov (2002) suggested that the pile capacity gain relationship given by

Equation (6.25a) should be used only as a guide for assessment of pile capacity with time.The pile capacity vs time relationships proposed by Skov and Denver (1989) and Svinkin andSkov (2002) indicate that when plotted on a log time scale the capacity gain continues

indefinitely Chun et al (1999) showed that the capacity gain was not infinite but eventuallyconverged to a constant value (long-term capacity) Chun et al (1999) observed that there is a

linear relationship between the ratio of bearing capacity gain (Ru(t) /REOID) and the rate of the

ratio of capacity gain d/dt[Ru(t)/REOID], where REOID and Ru(t) are EOID capacity and the

capacity at time t after installation, respectively, and proposed the following general

relationship to estimate the capacity gain regardless of the soil type:

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(6.26a)

where C=(Ru(∞)/REOID), which is the long-term ratio of capacity gain, B is a constant (=G/K),

K is the dissipation factor, and G is the aging factor.

Factor K depends on the permeability of the soil while factor G depends on the aging

potential associated with soil properties A higher B value could be expected for sandy soils, where the aging effect is predominant, and a lower value of B could be associated with clayey

soils as the pore pressure dissipation is predominant The ratio of capacity gain can be

obtained by solving the above differential equation:

(6.26b)

6.5.1 Reported Results from Field Studies

Tomlinson (1994) reports the results of pile load tests carried out on 200×215 mm piles intosoft clay at different times after installation Figure 6.13(a)shows the measured and estimated(Skov and Denver, 1989) capacity gain ratio for one pile;Figure 6.13(b)shows the measuredand estimated (Svinkin and Skov, 2002; Chun et al., 1999) capacity gain ratio for the samepile

It is evident fromFigures 6.13(a) that for large lapsed times after EOID the relationshipproposed by Chun et al (1999) predicts the capacity gain ratio over the entire time durationbetter than the method proposed by Svinkin et al (2002) for the case studies considered above

Two material parameters, B and C, are needed for the Chun method of capacity prediction where B and C are the long-term capacity gain ratio and a material constant, respectively The

relevant value of B is obtained by considering the time capacity variation of the measured

capacity gain ratio whereas parameter C is obtained by matching the measured with the

predicted values of the capacity gain ratio The values of parameters B and C estimated by

Thilakasiri et al (2003) and available values obtained from the literature are shown inTable6.8

Also indicated inTable 6.8are the times taken to develop 90% of the long-term capacityand the percentage of the long-term capacity developed 1 week after the EOID.Table 6.8shows that the 1-week wait period from the EOID is sufficient for piles in sand whereas the 1-week period is not enough for piles driven into clay deposits for which a minimum wait

period of 2 to 3 weeks may be required

More recently, Bullock et al (2005) published their test findings on a Florida test pileprogram The Florida DOT commonly uses 457mm (18 in.), square, prestressed, concretepiles to support low-level bridges Bullock et al (2005) provided the instrumentation andinstalled dedicated test piles of this type at four bridge construction sites in northern Florida.Each pile included an O-cell cast into the tip, strain gauges at soil layer boundaries, and totalstress cells and pore pressure cells centered in one pile face between adjacent strain gaugeelevations They calculated the shear force and average shear stress acting on the face of thepile from the difference in load between adjacent strain gauge levels The strain gauges

defined a total of 28 side shear segments, of which 18 also included pore pressure and total

expressed Equation (6.25a) as

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FIGURE 6.13

(a) Measured and estimated capacity from Svinkin et al (2002) and Chun et al (1999) for the pile I of

Tomlinson (1994) (From Svinkin, M.R and Skov, R., 2002, Setup effect of cohesive soils

in pile capacity, 2002, http://www.vulcanhammer.net/svinkin/set.htm With permission.) (b) Side shear setup results from the Florida pile testing program (From Bullock, P.J.,

Schmertmann, J.H., Mcvay, M.C., and Townsend, F., 2005, Journal of Geotechnical and Geoenvironmental Engineering, 131(3): 292–300 With permission.)

(1994)—pile II

Soft clay

Source: From Thilakasiri, H.S., Abeyasinghe, R.M., and Tennakoon, B.L., 2003, A study of strength gain of

driven piles, Proceedings of the 9th Annual Symposium, Engineering Research Unit, University of Moratuwa, Sri

Lanka With permission.

(6.25b)

where Q is the capacity of the entire pile in subsequent segmental analysis, Q0is the capacity

at initial reference time t0, t is the time since EOID, t0is the reference time since EOID, and A

is the dimensionless setup factor

Bullock et al (2005) presented the following relationship between the segmental side shearsetup factors and the side shear setup factor for the entire pile:

(6.25c)

where fs0iis the unit side shear stress at time t0for segment i, Liis the length of segment i, and

A i is the side shear setup factor for segment i.

Values of A obtained by Bullock et al (2005) for staged and unstaged tests are shown in

Figure 6.13(b) Bullock et al (2005) recommend a reduction factor of Cst=(Aunstaged/

Astaged)=0.4 for all soil types to correct setup A factors measured using staged field tests,

including repeated dynamic re-strikes, repeated static tests, or repeated SPT-Ts