Seismic Design Practice in Japan doc

Seismic Design Practice in Japan 44.1 Introduction44.2 History of Earthquake Damage and Development of Seismic Design Methods44.3 Damage of Highway Bridges Caused by the Hyogo-ken Nanbu

Unjoh, S "Seismic Design Practice in Japan."

Bridge Engineering Handbook

Ed Wai-Fah Chen and Lian Duan

Boca Raton: CRC Press, 2000

Seismic Design Practice

in Japan

44.1 Introduction44.2 History of Earthquake Damage and Development of Seismic Design Methods44.3 Damage of Highway Bridges Caused by the Hyogo-ken Nanbu Earthquake

44.4 1996 Seismic Design Specifications

Foundations • Design against Soil Liquefaction and Liquefaction-induced Lateral Spreading • Bearing Supports • Unseating Prevention Systems

44.5 Seismic Retrofit Practices for Highway Bridges

Past Seismic Retrofit Practices • Seismic Retrofit after the Hyogo-ken Nanbu Earthquake

Nomenclature

The following symbols are used in this chapter The section number in parentheses after definition

of a symbol refers to the section where the symbol first appears or is defined

a space of tie reinforcement (Section 44.4.4)

ACF sectional area of carbon fiber (Figure 44.19)

A h area of tie reinforcements (Section 44.4.4)

A w sectional area of tie reinforcement (Section 44.4.4)

b width of section (Section 44.4.4)

c B coefficient to evaluate effective displacement (Section 44.4.7)

c B modification coefficient for clearance (Section 44.4.11)

c df modification coefficient (Section 44.4.2)

c c modification factor for cyclic loading (Section 44.4.4)

c D modification coefficient for damping ratio (Section 44.4.6)

c e modification factor for scale effect of effective width (Section 44.4.4)

c E modification coefficient for energy-dissipating capability (Section 44.4.7)

c P coefficient depending on the type of failure mode (Section 44.4.2)

c pt modification factor for longitudinal reinforcement ratio (Section 44.4.4)Shigeki Unjoh

Public Works Research Institute

c R factor depending on the bilinear factor r (Section 44.4.2)

c W corrective coefficient for ground motion characteristics (Section 44.4.9)

c Z modification coefficient for zone (Section 44.4.3)

d effective width of tie reinforcements (Section 44.4.4)

d height of section (Section 44.4.4)

D a width or a diameter of a pier (Section 44.4.4)

D E coefficient to reduce soil constants according to F L value (Section 44.4.11)

E c elastic modules of concrete (Section 44.4.4)

E CF elastic modulus of carbon fiber (Figure 44.19)

E des gradient at descending branch (Section 44.4.4)

F L liquefaction resistant ratio (Section 44.4.9)

F(u) restoring force of a device at a displacement u (Section 44.4.7)

h height of a pier (Section 44.4.4)

h B height of the center of gravity of girder from the top of bearing (Figure 44.13)

h B equivalent damping of a Menshin device (Section 44.4.7)

h i damping ratio of ith mode (Section 44.4.6)

h ij damping ratio of jth substructure in ith mode (Section 44.4.6)

h Bi damping ratio of ith damper (Section 44.4.7)

h Pi damping ratio of ith pier or abutment (Section 44.4.7)

h Fui damping ratio of ith foundation associated with translational displacement (Section 44.4.7)

h Fθi damping ratio of ith foundation associated with rotational displacement(Section 44.4.7)

H distance from a bottom of pier to a gravity center of a deck (Section 44.4.7)

H0 shear force at the bottom of footing (Figure 44.12)

I importance factor (Section 44.5.2)

k hc lateral force coefficient (Section 44.4.2)

k hc design seismic coefficient for the evaluation of liquefaction potential (Section 44.4.9)

k hc0 standard modification coefficient (Section 44.4.3)

k hcm lateral force coefficient in Menshin design (Section 44.4.7)

k he equivalent lateral force coefficient (Section 44.4.2)

k hem equivalent lateral force coefficient in Menshin design (Section 44.4.7)

k hp lateral force coefficient for a foundation (Section 44.4.2)

k j stiffness matrix of jth substructure (Section 44.4.6)

K stiffness matrix of a bridge (Section 44.4.6)

K B equivalent stiffness of a Menshin device (Section 44.4.7)

K Pi equivalent stiffness of ith pier or abutment (Section 44.4.7)

K Fui translational stiffness of ith foundation (Section 44.4.7)

K Fθi rotational stiffness of ith foundation (Section 44.4.7)

L shear stress ratio during an earthquake (Section 44.4.9)

L A redundancy of a clearance (Section 44.4.11)

L E clearance at an expansion joint (Section 44.4.11)

L P plastic hinge length of a pier (Section 44.4.4)

M0 moment at the bottom of footing (Figure 44.12)

P a lateral capacity of a pier (Section 44.4.2)

P s shear capacity in consideration of the effect of cyclic loading (Section 44.4.4)

P s0 shear capacity without consideration of the effect of cyclic loading (Section 44.4.4)

P u bending capacity (Section 44.4.2)

r bilinear factor defined as a ratio between the first stiffness (yield stiffness) and the second

stiffness (postyield stiffness) of a pier (Section 44.4.2)

r d modification factor of shear stress ratio with depth (Section 44.4.9)

R dynamic shear strength ratio (Section 44.4.9)

R priority (Section 44.5.2)

R D dead load of superstructure (Section 44.4.11)

R heq and R veq vertical reactions caused by the horizontal seismic force and vertical force (Section 44.4.11)

R L cyclic triaxial strength ratio (Section 44.4.9)

R U design uplift force applied to the bearing support (Section 44.4.11)

s space of tie reinforcements (Section 44.4.4)

S earthquake force (Section 44.5.2)

S c shear capacity shared by concrete (Section 44.4.4)

SI and SII acceleration response spectrum for Type-I and Type-II ground motions (Section 44.4.6)

S I0 and S II0standard acceleration response spectrum for Type-I and Type-II ground motions

(Section 44.4.6)

S E seat length (Section 44.4.11)

S EM minimum seat length (cm) (Section 44.4.11)

S s shear capacity shared by tie reinforcements (Section 44.4.4)

T natural period of fundamental mode (Table 44.3)

∆T difference of natural periods (Section 44.4.11)

T1 and T2 natural periods of the two adjacent bridge systems (Section 44.4.11)

u B design displacement of isolators (Section 44.4.7)

u Be effective design displacement (Section 44.4.7)

u Bi design displacement of ith Menshin device (Section 44.4.7)

u G relative displacement of ground along the bridge axis (Section 44.4.11)

u R relative displacement (cm) developed between a superstructure and a substructure

(Section 44.4.11)

V0 vertical force at the bottom of footing (Figure 44.12)

V T structural factor (Section 44.5.2)

V RP1 design specification (Section 44.5.2)

V RP2 pier structural factor (Section 44.5.2)

V RP3 aspect ratio (Section 44.5.2)

V MP steel pier factor (Section 44.5.2)

V FS unseating device factor (Section 44.5.2)

V F foundation factor (Section 44.5.2)

w v weighting factor on structural members (Section 44.5.2)

W equivalent weight (Section 44.4.2)

W elastic strain energy (Section 44.4.7)

W P weight of a pier (Section 44.4.2)

W U weight of a part of superstructure supported by the pier (Section 44.4.2)

∆W energy dissipated per cycle (Section 44.4.7)

α safety factor (Section 44.4.4)

α, β coefficients depending on shape of pier (Section 44.4.4)

αm safety factor used in Menshin design (Section 44.4.7)

δy yield displacement of a pier (Section 44.4.2)

δR residual displacement of a pier after an earthquake (Section 44.4.2)

δRa allowable residual displacement of a pier (Section 44.4.2)

δu ultimate displacement of a pier (Section 44.4.4)

εc strain of concrete (Section 44.4.4)

εcc strain at maximum strength (Section 44.4.4)

εG ground strain induced during an earthquake along the bridge axis (Section 44.4.11)

εs strain of reinforcements (Section 44.4.4)

εsy yield strain of reinforcements (Section 44.4.4)

θ angle between vertical axis and tie reinforcement (Section 44.4.4)

θpu ultimate plastic angle (Section 44.4.4)

µ a allowable displacement ductility factor of a pier (Section 44.4.2)

µ m allowable ductility factor of a pier in Menshin design (Section 44.4.7)

µ R response ductility factor of a pier (Section 44.4.2)

ρs tie reinforcement ratio (Section 44.4.4)

σc stress of concrete (Section 44.4.4)

σcc strength of confined concrete (Section 44.4.4)

σCF stress of carbon fiber (Figure 44.19)

σck design strength of concrete (Section 44.4.4)

σs stress of reinforcements (Section 44.4.4)

σsy yield strength of reinforcements (Section 44.4.4)

σv total loading pressure (Section 44.4.9)

σ′v effective loading pressure (Section 44.4.9)

τc shear stress capacity shared by concrete (Section 44.4.4)

φij mode vector of jth substructure in ith mode (Section 44.4.6)

φi mode vector of a bridge in ith mode (Section 44.4.6)

φy yield curvature of a pier at bottom (Section 44.4.4)

φu ultimate curvature of a pier at bottom (Section 44.4.4)

44.1 Introduction

Japan is one of the most seismically disastrous countries in the world and has often sufferedsignificant damage from large earthquakes More than 3000 highway bridges have suffered damagesince the 1923 Kanto earthquake The earthquake disaster prevention technology for highway bridgeshas been developed based on such bitter damage experiences Various provisions for designingbridges have been developed to prevent damage due to the instability of soils such as soil liquefaction.Furthermore, design detailings including unseating prevention devices are implemented Withprogress in improving seismic design provisions, damage to highway bridges caused by the earth-quakes has been decreasing in recent years

However, the Hyogo-ken Nanbu earthquake of January 17, 1995 caused destructive damage tohighway bridges Collapse and near collapse of superstructures occurred at nine sites, and otherdestructive damage occurred at 16 sites [1] The earthquake revealed that there are a number ofcritical issues to be revised in the seismic design and seismic retrofit of bridges [2,3]

This chapter presents technical developments for seismic design and seismic retrofit of highwaybridges in Japan The history of the earthquake damage and development of the seismic designmethods is first described The damage caused by the 1995 Hyogo-ken Nanbu earthquake, the

lessons learned from the earthquake, and the seismic design methods introduced in the 1996 Seismic

Design Specifications for Highway Bridges are then described Seismic performance levels and design

methods as well as ductility design methods for reinforced concrete piers, steel piers, foundations,and bearings are described Then the history of the past seismic retrofit practices is described Theseismic retrofit program after the Hyogo-ken-Nanbu earthquake is described with emphasis on theseismic retrofit of reinforced concrete piers as well as research and development on the seismicretrofit of existing highway bridges

44.2 History of Earthquake Damage and Development of Seismic

of provisions in seismic design for highway bridges

In particular, the seismic design method was integrated and upgraded by compiling the fications for Seismic Design of Highway Bridges” in 1971 The design method for soil liquefactionand unseating prevention devices was introduced in the Specifications It was revised in 1980 andintegrated as “Part V: Seismic Design” in Design Specifications of Highway Bridges The primitivecheck method for ductility of reinforced concrete piers was included in the reference of the Speci-fications It was further revised in 1990 and ductility check of reinforced concrete piers, soil lique-faction, dynamic response analysis, and design detailings were prescribed It should be noted herethat the detailed ductility check method for reinforced concrete piers was first introduced in the

“Speci-1990 Specifications

However, the Hyogo-ken Nanbu earthquake of January 17, 1995, exactly 1 year after theNorthridge earthquake, California, caused destructive damage to highway bridges as describedearlier After the earthquake the Committee for Investigation on the Damage of Highway BridgesCaused by the Hyogo-ken Nanbu Earthquake (chairman, Toshio Iwasaki, Executive Director, CivilEngineering Research Laboratory) was established in the Ministry of Construction to investigatethe damage and to identify the factors that caused the damage

TABLE 44.1 History of Seismic Design Methods

1926 Details

of Road Structure (draft) Road Law, MIA

1939 Design Specifications

of Steel Highway Bridges (draft) MIA

1956 Design Specifications

of Steel Highway Bridges, MOC

1964 Design Specifications

of Substructures (Pile Foundations), MOC

1964 Design Specifications

of Steel Highway Bridges, MOC

1966 Design Specifications

of Substructures (Survey and Design), MOC

1968 Design Specifications

of Substructures (Piers and Direct Foundations), MOC

1970 Design Specifications

of Substructures (Caisson Foundations), MOC

1971 Specifications for Seismic Design

of Highway Bridges, MOC

1972 Design Specifications

of Substructures (Cast-in-Piles), MOC

1975 Design Specifications

of Substructures (Pile Foundations), MOC

1980 Design Specifications

of Highway Bridges, MOC

1990 Design Specifications

of Highway Bridges, MOC Seismic

k h = 0.1–0.3 Integration of seismic coefficient method and modified one.

section area such as RC frame and hollow section

Check of shear strength

Provision of definite design method, decreasing of allowable shear stress

Bearing capacity for

capacity for lateral force

Pile

foundation Bearing capacity in vertical direction was supposed to be checked Provisions of Definite Design Method (bearing capacity in vertical and horizontal directions) Special Condition (Foundation on Slope, Provisions of Design Details for Pile Head

Consolidation Settlement, Lateral Movement) Direct

foundation Supposed to be designed in a similar way provided in Design Specification of Caisson Foundation of 1969 Provisions of Definite Design Method

Soil

Liquefaction

Provisions of soil layers of which bearing capacity shall be ignored in seismic design

Provisions of evaluation method

of soil liquefaction and the treatment in seismic design

Consideration

of effect of fine sand content Bearing

support

supports (bearing, roller, anchor bolt)

Provision of transmitting method of seismic load at bearing Devices preventing

falling-off of

superstructure

Provision of bearing seat

length S

Provisions of stopper at movable bearings, devices for preventing superstructure from falling (seat

length S, connection of adjacent decks)

Provisions of stopper at movable bearings, devices for preventing superstructure from falling (seat

length Sε devices)

© 2000 by CRC Press LLC

On February 27, 1995, the Committee approved the “Guide Specifications for Reconstructionand Repair of Highway Bridges Which Suffered Damage Due to the Hyogo-ken Nanbe Earthquake,”

[4], and the Ministry of Construction announced on the same day that the reconstruction andrepair of the highway bridges which suffered damage in the Hyogo-ken Nanbu earthquake should

be made by the Guide Specifications It was decided by the Ministry of Construction on May 25,

1995 that the Guide Specifications should be tentatively used in all sections of Japan as emergencymeasures for seismic design of new highway bridges and seismic strengthening of existing highwaybridges until the Design Specifications of Highway Bridges is revised

In May, 1995, the Special Sub-Committee for Seismic Countermeasures for Highway Bridges(chairman, Kazuhiko Kawashima, Professor of the Tokyo Institute of Technology) was established

in the Bridge Committee (chairman, Nobuyuki Narita, Professor of the Tokyo Metropolitan versity), Japan Road Association, to draft the revision of the Design Specifications of HighwayBridges The new Design Specifications of Highway Bridges [5,6] was approved by the BridgeCommittee, and issued by the Ministry of Construction on November 1, 1996

Uni-44.3 Damage of Highway Bridges Caused by the Hyogo-ken

Nanbu Earthquake

The Hyogo-ken Nanbu earthquake was the first earthquake to hit an urban area in Japan since the

1948 Fukui earthquake Although the magnitude of the earthquake was moderate (M7.2), theground motion was much larger than anticipated in the codes It occurred very close to Kobe Citywith shallow focal depth

Damage was developed at highway bridges on Routes 2, 43, 171, and 176 of the National Highway,Route 3 (Kobe Line) and Route 5 (Bay Shore Line) of the Hanshin Expressway, and the Meishinand Chugoku Expressways Damage was investigated for all bridges on national highways, theHanshin Expressway, and expressways in the area where destructive damage occurred The totalnumber of piers surveyed reached 3396 [1] Figure 44.1 shows Design Specifications referred to inthe design of the 3396 highway bridges Most of the bridges that suffered damage were designedaccording to the 1964 Design Specifications or the older Design Specifications Although the seismicdesign methods have been improved and amended several times since 1926, only a requirement forlateral force coefficient was provided in the 1964 Design Specifications or the older Specifications

Figure 44.2 compares damage of piers (bridges) on the Route 3 (Kobe Line) and Route 5 (BayShore Line) of the Hanshin Expressway Damage degree was classified as As (collapse), A (nearlycollapse), B (moderate damage), C (damage of secondary members), and D (minor or no damage).Substructures on Route 3 and Route 5 were designed with the 1964 Design Specifications and the

1980 Design Specifications, respectively It should be noted in this comparison that the intensity of

FIGURE 44.1 Design specifications referred to in design of Hanshin Expressway [2].

ground shaking in terms of response spectra was smaller at the Bay Area than the narrow rectangulararea where JMA seismic intensity was VII (equivalent to modified Mercalli intensity of X-XI) Route

3 was located in the narrow rectangular area, while Route 5 was located in the Bay Area Keeping

in mind such differences in ground motion, it is apparent in Figure 44.2 that about 14% of the piers

on Route 3 suffered As or A damage while no such damage was developed in the piers on Route 5.Although damage concentrated on the bridges designed with the older Design Specifications, itwas thought that essential revision was required even in the recent Design Specifications to preventdamage against destructive earthquakes such as the Hyogo-ken Nanbu earthquake The main mod-ifications were as follows:

1 To increase lateral capacity and ductility of all structural components in which seismic force

is predominant so that ductility of a total bridge system is enhanced For such purpose, itwas required to upgrade the “Check of Ductility of Reinforced Concrete Piers,” which hasbeen used since 1990, to a “ductility design method” and to apply the ductility design method

to all structural components It should be noted here that “check” and “design” are different;the check is only to verify the safety of a structural member designed by another designmethod, and is effective only to increase the size or reinforcements if required, while thedesign is an essential procedure to determine the size and reinforcements

2 To include the ground motion developed at Kobe in the earthquake as a design force in theductility design method

3 To specify input ground motions in terms of acceleration response spectra for dynamicresponse analysis more actively

4 To increase tie reinforcements and to introduce intermediate ties for increasing ductility ofpiers It was decided not to terminate longitudinal reinforcements at midheight to preventpremature shear failure, in principle

5 To adopt multispan continuous bridges for increasing number of indeterminate of a totalbridge system

6 To adopt rubber bearings for absorbing lateral displacement between a superstructure andsubstructures and to consider correct mechanism of force transfer from a superstructure tosubstructures

7 To include the Menshin design (seismic isolation)

8 To increase strength, ductility, and energy dissipation capacity of unseating preventiondevices

9 To consider the effect of lateral spreading associated with soil liquefaction in design offoundations at sites vulnerable to lateral spreading

FIGURE 44.2 Comparison of damage degree between Route 3 (a) and Route 5 (b) (As: collapse, A: near collapse,

B: moderate damage, C: damage of secondary members, D: minor or no damage) [2]

44.4 1996 Seismic Design Specifications of Highway Bridges

44.4.1 Basic Principles of Seismic Design

The 1995 Hyogo-ken Nanbu earthquake, the first earthquake to be considered that such destructivedamage could be prevented due to the progress of construction technology in recent years, provided

a large impact on the earthquake disaster prevention measures in various fields Part V: SeismicDesign of the Design Specifications of Highway Bridges (Japan Road Association) was totally revised

in 1996, and the design procedure moved from the traditional seismic coefficient method to theductility design method The revision was so comprehensive that the past revisions of the last 30years look minor

A major revision of the 1996 Specifications is the introduction of explicit two-level seismic designconsisting of the seismic coefficient method and the ductility design method Because Type I and Type IIground motions are considered in the ductility design method, three design seismic forces are used indesign Seismic performance for each design force is clearly defined in the Specifications

Table 44.2 shows the seismic performance level provided in the 1996 Design Specifications Thebridges are categorized into two groups depending on their importance: standard bridges (Type Abridges) and important bridges (Type B bridges) The seismic performance level depends on theimportance of the bridge For moderate ground motions induced in earthquakes with a highprobability of occurrence, both A and B bridges should behave in an elastic manner without essentialstructural damage For extreme ground motions induced in earthquakes with a low probability ofoccurrence, Type A bridges should prevent critical failure, whereas Type B bridges should performwith limited damage

In the ductility design method, two types of ground motions must be considered The first is theground motions that could be induced in plate boundary-type earthquakes with a magnitude of about

8 The ground motion at Tokyo in the 1923 Kanto earthquake is a typical target of this type of groundmotion The second is the ground motion developed in earthquakes with magnitude of about 7 to 7.2

at very short distance Obviously, the ground motions at Kobe in the Hyogo-ken Nanbu earthquake is

a typical target of this type of ground motion The first and the second ground motions are calledType I and Type II ground motions, respectively The recurrence time of Type II ground motion may

be longer than that of Type I ground motion, although the estimation is very difficult

The fact that lack of near-field strong motion records prevented serious evaluation of the validity

of recent seismic design codes is important The Hyogo-ken Nanbu earthquake revealed that thehistory of strong motion recording is very short, and that no near-field records have yet beenmeasured by an earthquake with a magnitude on the order of 8 It is therefore essential to havesufficient redundancy and ductility in a total bridge system

TABLE 44.2 Seismic Performance Levels

Type of Design Ground Motions

Importance of Bridges Design Methods Type-A

(Standard Bridges)

Type-B (Important Bridges)

Equivalent Static Lateral Force Methods

Dynamic Analysis Ground motions with high

probability to occur

Prevent Damage Seismic

coefficient method

Step by Step analysis or Response spectrum analysis

Ground motions with low

probability to occur

Type I (plate boundary earthquakes)

Prevent critical damage

Limited damage

Ductility design method Type II

(Inland earthquakes)

In the ductility design method, assuming a principal plastic hinge is formed at the bottom ofpier as shown in Figure 44.4a and that the equal energy principle applies, a bridge is designed sothat the following requirement is satisfied:

weight of a part of superstructure supported by the pier, W P = weight of a pier, and c P = coefficient

depending on the type of failure mode The c P is 0.5 for a pier in which either flexural failure orshear failure after flexural cracks are developed, and 1.0 is for a pier in which shear failure is

developed The lateral capacity of a pier P a is defined as a lateral force at the gravity center of asuperstructure

In Type B bridges, residual displacement developed at a pier after an earthquake must be checkedas

in which δR = residual displacement of a pier after an earthquake, δRa = allowable residual

displace-ment of a pier, r = bilinear factor defined as a ratio between the first stiffness (yield stiffness) and the second stiffness (postyield stiffness) of a pier, c R = factor depending on the bilinear factor r, µ R =response ductility factor of a pier, and δy = yield displacement of a pier The δRa should be 1/100 ofthe distance between the bottom of a pier and the gravity center of a superstructure

In a bridge with complex dynamic response, the dynamic response analysis is required to checkthe safety of a bridge after it is designed by the seismic coefficient method and the ductility designmethod Because this is only for a check of the design, the size and reinforcements of structuralmembers once determined by the seismic coefficient method and the ductility design methods may

be increased if necessary It should be noted, however, that under the following conditions in whichthe ductility design method is not directly applied, the size and reinforcements can be determinedbased on the results of a dynamic response analysis as shown in Figure 44.3 Situations when theductility design method should not be directly used include:

1 When principal mode shapes that contribute to bridge response are different from the onesassumed in the ductility design methods

2 When more than two modes significantly contribute to bridge response

3 When principal plastic hinges form at more than two locations, or principal plastic hingesare not known where to be formed

4 When there are response modes for which the equal energy principle is not applied

In the seismic design of a foundation, a lateral force equivalent to the ultimate lateral capacity

of a pier P u is assumed to be a design force as

(44.7)

in which k hp = lateral force coefficient for a foundation, c df = modification coefficient (= 1.1), and

W = equivalent weight by Eq (44.3) Because the lateral capacity of a wall-type pier is very large in

the transverse direction, the lateral seismic force evaluated by Eq (44.7) in most cases becomesexcessive Therefore, if a foundation has sufficiently large lateral capacity compared with the lateralseismic force, the foundation is designed assuming a plastic hinge at the foundation and surroundingsoils as shown in Figure 44.4c

44.4.3 Design Seismic Force

Lateral force coefficient k hc in Eq (44.2) is given as

(44.8)

in which c Z = modification coefficient for zone, and is 0.7, 0.85, and l.0 depending on the zone, and

k hc0 = standard modification coefficient Table 44.3 and Figure 44.5 show the standard lateral force

coefficients k hc0 for Type I and Type II ground motions Type I ground motions have been used since

1990 (1990 Specifications), while Type II ground motions were newly introduced in the 1996

Specifications It should be noted here that the k hc0 at stiff site (Group I) has been assumed smallerthan the khc0 at moderate (Group II) and soft soil (Group III) sites in Type I ground motions aswell as the seismic coefficients used for the seismic coefficient method Type I ground motions wereessentially estimated from an attenuation equation for response spectra that is derived from astatistical analysis of 394 components of strong motion records Although the response spectralaccelerations at short natural period are larger at stiff sites than at soft soil sites, the tendency hasnot been explicitly included in the past This was because damage has been more developed at softsites than at stiff sites To consider such a fact, the design force at stiff sites is assumed smaller than

k hp=c P W df u

k hc= ⋅c k z hc0

that at soft sites even at short natural period However, being different from such a traditionalconsideration, Type II ground motions were determined by simply taking envelopes of responseaccelerations of major strong motions recorded at Kobe in the Hyogo-ken Nanbu earthquake.Although the acceleration response spectral intensity at short natural period is higher in Type IIground motions than in Type I ground motions, the duration of extreme accelerations excursion

is longer in Type I ground motions than Type II ground motions As will be described later, such

a difference of the duration has been taken into account to evaluate the allowable displacementductility factor of a pier

44.4.4 Ductility Design of Reinforced Concrete Piers

44.4.4.1 Evaluation of Failure Mode

In the ductility design of reinforced concrete piers, the failure mode of the pier is evaluated as thefirst step Failure modes are categorized into three types based on the flexural and shear capacities

of the pier as

1 P u  Ps bending failure

2 P s ≤ P u  P s0 bending to shear failure

3 P s0 ≤ P u shear failure

TABLE 44.3 Lateral Force Coefficient k hc0 in the Ductility Design Method

Soil Condition Lateral Force Coefficient k hc0

Type I Ground Motion Group I (stiff) k hc0 = 0.7 for T  1.4 k hc0 = 0.876T 2/3 for T > 1.4

Group II (moderate) k hc0 =1.51T 1/3 (k hc0  0.7) for T <0.18 k hc0 = 0.85 for 0.18  T  1.6 khc0 = 1.16T 2/3 for T > 1.6

Group III (soft) k hc0 = 1.51T 1/3 (k hc0  0.7) for T < 0.29 khc0 = 1.0 for 0.29  T  2.0 k hc0 = 1.59T 2/3 for T >2.0

Type II Ground Motion Group I (stiff) k hc0 = 4.46T 2/3 for T  0.3 k hc0 = 2.00 for 0.3  T  0.7 k hc0 = 1.24T 4/3 for T > 0.7

Group II (moderate) k hc0 = 3.22T 2/3 for T < 0.4 k hc0 = 1.75 for 0.4  T  1.2 k hc0 =2.23T 4/3 for T > 1.2

Group III (soft) k hc0 =2.38T 2/3 for T < 0.5 k hc0 = 1.50 for 0.5  T  1.5 khc0 = 2.57T 4/3 for T > 1.5

FIGURE 44.5 Type I and Type II ground motions in the ductility design method.

in which P u = bending capacity, P s = shear capacity in consideration of the effect of cyclic loading,

and P s0 = shear capacity without consideration of the effect of cyclic loading

The ductility factor and capacity of the reinforced concrete piers are determined according to thefailure mode as described later

44.4.4.2 Displacement Ductility Factor

Th allowable displacement ductility factor of a pier µa in Eq (44.2) is evaluated as

(44.9)

in which α = safety factor, δy = yield displacement of a pier, and δu = ultimate displacement of a

pier As well as the lateral capacity of a pier P a in Eq (44.1), the δy and δ u are defined at the gravitycenter of a superstructure In a reinforced concrete single pier as shown in Figure 44.4a, the ultimatedisplacement δu is evaluated as

(44.10)

in which φy = yield curvature of a pier at bottom, φu = ultimate curvature of a pier at bottom, h = height of a pier, and L P = plastic hinge length of a pier The plastic hinge length is given as

(44.11)

in which D is a width or a diameter of a pier.

The yield curvature φy and ultimate curvature φu in Eq (44.10) are evaluated assuming astress–strain relation of reinforcements and concrete as shown in Figure 44.6 The stress σc – strain

εc relation of concrete with lateral confinement is assumed as

(44.12)

(44.13)

in which σcc = strength of confined concrete, E c = elastic modules of concrete, εcc = strain at

max-imum strength, and E des = gradient at descending branch In Eq (44.12), σcc, εcc , and E des aredetermined as

in which σck = design strength of concrete, σsy = yield strength of reinforcements, α and β =coefficients depending on shape of pier (α = 1.0 and β = 1.0 for a circular pier, and α = 0.2 and

β = 0.4 for a rectangular pier), and ρs = tie reinforcement ratio defined as

A sd

E

=+

it is known that a certain level of failure in a pier such as a sudden decrease of lateral capacity occurs

at smaller lateral displacement in a pier subjected to a loading hysteresis with a greater number ofload reversals To reflect such a fact, it was decided that the ultimate strain εcu should be evaluated

by Eq (44.18), depending on the type of ground motions Therefore, the allowable ductility factor µadepends on the type of ground motions; the µa is larger in a pier subjected to Type II ground motionsthan a pier subjected to Type I ground motions

It should be noted that the safety factor α in Eq (44.9) depends on the type of bridges as well

as the type of ground motions as shown in Table 44.4 This is to preserve higher seismic safety inthe important bridges, and to take account of the difference of recurrent time between Type I andType II ground motions

44.4.4.3 Shear Capacity

Shear capacity of reinforced concrete piers is evaluated by a conventional method as

(44.19)(44.20)

(44.21)

in which P s = shear capacity; S c = shear capacity shared by concrete; S s = shear capacity shared bytie reinforcements, τc = shear stress capacity shared by concrete; c c = modification factor for cyclic

loading (0.6 for Type I ground motions; 0.8 for Type II ground motions); c e = modification factor

for scale effect of effective width; c pt = modification factor for longitudinal reinforcement ratio; b and d = width and height of section, A w = sectional area of tie reinforcement; σsy = yield strength

of tie reinforcement, θ = angle between vertical axis and tie reinforcement, and a = space of tie

reinforcement

The modification factor on the scale effect of effective width, c e, was based on experimental study

of loading tests of beams with various effective heights and was newly introduced in the 1996Specifications Table 44.5 shows the modification factor on scale effect

44.4.4.4 Arrangement of Reinforcement

Figure 44.7 shows a suggested arrangement of tie reinforcement Tie reinforcement should bedeformed bars with a diameter equal or larger than 13 mm, and it should be placed in most bridges

TABLE 44.4 Safety Factor α in Eq 44.9

Type of Bridges Type I Ground Motion Type II Ground Motion Type B 3.0 1.5

Type A 2.4 1.2

TABLE 44.5 Modification Factor on Scale

Effect for Shear Capacity Shared by Concrete

Effective Width of Section d (m) Coefficient c c

at a distance of no longer than 150 mm In special cases, such as bridges with pier height taller than

30 m, the distance of tie reinforcement may be increased at height so that pier strength should not

be sharply decreased at the section Intermediate ties should be also provided with the same distancewith the ties to confine the concrete Space of the intermediate ties should be less than 1 m

where the plastic hinge length L P is assumed to be Eq (44.11)

When the two-column bent is subjected to lateral force in the transverse direction, axial forcedeveloped in the beam and columns is affected by the applied lateral force Therefore, the horizontalforce–displacement relation is obtained through the static push-over analysis considering axial force

N/moment M interaction relation The ultimate state of each plastic hinge is obtained by the ultimate

plastic angle θpu as

(44.22)

in which φu = ultimate curvature and φy = yield curvature

The ultimate state of the whole two-bent column is determined so that all four plastic hingesdeveloped reach the ultimate plastic angle

FIGURE 44.7 Confinement of core concrete by tie reinforcement (a) Square section; (b) semisquare section; (c)

circular section; (d) hollow section.

θpu=(φ φu y−1)L Pφy

44.4.5 Ductility Design of Steel Piers

44.4.5.1 Basic Concept

To improve seismic performance of a steel pier, it is important to avoid specific brittle failure modes

Figure 44.9 shows the typical brittle failure mode for rectangular and circular steel piers Thefollowing are the countermeasures to avoid such brittle failure modes and to improve seismicperformance of steel piers:

1 Fill the steel column with concrete

2 Improve structural parameters related to buckling strength

• Decrease the width–thickness ratio of stiffened plates of rectangular piers or the ter–thickness ratio of steel pipes;

diame-• Increase the stiffness of stiffeners;

• Reduce the diaphragm spacing;

• Strengthen corners using the corner plates;

3 Improve welding section at the corners of rectangular section

4 Eliminate welding section at the corners by using round corners

FIGURE 44.8 Analytical idealization of a two-column bent.

FIGURE 44.9 Typical brittle failure modes of steel piers (a) Fracture of corners; (b) elephant knee buckling.

44.4.5.2 Concrete-Infilled Steel Pier

In a concrete-infilled steel pier, the lateral capacity P a and the allowable displacement ductilityfactor µa in Eqs (44.1) and (44.2) are evaluated as

(44.23)

(44.24)

in which P y and P u = yield and ultimate lateral capacity of a pier; δy and δu = yield and ultimatedisplacement of a pier; and α = safety factor (refer to Table 44.4) The P a and the µa are evaluatedidealizing that a concrete-infilled steel pier resists flexural moment and shear force as a reinforcedconcrete pier It is assumed in this evaluation that the steel section is idealized as reinforcing barsand that only the steel section resists axial force A stress vs strain relation of steel and concrete asshown in Figure 44.10 is assumed The height of infilled concrete has to be decided so that bucking

is not developed above the infilled concrete

44.4.5.3 Steel Pier without Infilled Concrete

A steel pier without infilled concrete must be designed with dynamic response analysis Properties

of the pier need to be decided based on a cyclic loading test Arrangement of stiffness and welding

at corners must be precisely evaluated so that brittle failure is avoided

44.4.6 Dynamic Response Analysis

Dynamic response analysis is required in bridges with complex dynamic response to check the safetyfactor of the static design Dynamic response analysis is also required as a “design” tool in the bridgesfor which the ductility design method is not directly applied In dynamic response analysis, groundmotions which are spectral-fitted to the following response spectra are used;

(44.25)(44.26)

in which SI and SII = acceleration response spectrum for Type I and Type II ground motions; SI 0and SII 0 = standard acceleration response spectrum for Type I and Type II ground motions, respec-

tively; c Z = modification coefficient for zone, refer to Eq (44.8); and c D = modification coefficientfor damping ratio given as

(44.27)

Table 44.6 and Figure 44.11 show the standard acceleration response spectra (damping ratio h =0.05) for Type I and Type II ground motions

It is recommended that at least three ground motions be used per analysis and that an average

be taken to evaluate the response

In dynamic analysis, modal damping ratios should be carefully evaluated To determine the modaldamping ratios, a bridge may be divided into several substructures in which the energy-dissipatingmechanism is essentially the same If one can specify a damping ratio of each substructure for a

given mode shape, the modal damping ratio for the ith mode, h i, may be evaluated as

P P

D i

=+ +

1 5