Advances in Steel Structures - part 10 ppt

Cantilever Columns with Pipe Sections The behavior of thin-walled steel cantilever-typed columns with pipe section has been investigated by some researchers e.g., Nishikawa et al.. Thus

70 T Usami et al

TABLE 3 PARAMETERS OF CANTILEVER COLUMNS WITH UNSTIFFENED BOX SECTIONS Specimen

UU1

UU6

UUll

U45-2513]

U45-4013]

U70-2513]

U70-4013]

(mm) (mm) (mm) (mm) (mm) k RI P/PY Number

762 306 1 5 7 5 120.0 4.51 0.362 0.664 0.2 I

1035 394 202.5 154.0 4.51 0.381 0.854 0.2 I

853 312 1 7 1 5 127.0 10.5 0.406 0.297 0.2 I

485 278 144.0 108.8 5 9 1 0.254 0.448 0.2 II

781 278 145.0 1 0 8 8 5.91 0.404 0.451 0.2 II

786 434 222.0 167.8 5.91 0.262 0.701 0.2 II

1217 434 222.9 167.8 5 9 1 0.406 0.704 0.2 II Notes: refer to Table 5 for details of material numbers

TABLE 4 PARAMETERS OF CANTILEVER COLUMNS WITH STIFFENED BOX SECTIONS Model

B1

B2

B3

B4

B5

B6

B7

B8

B14

B16

h

(mm)

4311

7543

7559

10776

3264

5712

5551

5777

3403

5712

(mm) (mm) (mm) (mm) RI A, A u P/P,

1344 20 121 20 0.46 0.51 0.20 1.0 0.15

1344 20 121 20 0.46 0 5 1 0.35 1.0 0.15

1344 20 121 20 0.46 0.28 0.35 0.5 0.15

1344 20 121 20 0.46 0.51 0.50 1.0 0.15

1023 20 105 20 0.35 0.21 0.20 0.5 0.15

1023 20 105 20 0.35 0.21 0.35 0.5 0.15

1023 20 179 20 0.35 0.23 0.35 1.0 0.15

1023 20 70 20 0.35 0.33 0.35 0.5 0.15

882 9 80 6 0.56 0.63 0.26 1.0 0.12

1023 20 105 20 0.35 0 4 1 0.35 1.0 0.15 Notes" D = B; refer to Table 5 for details of material numbers

Material Number III III III III III III III III

IV III

members Besides, the computed results are compared with those reported in previous studies (Usami 1996; Nishikawa, et al 1996; Gao 1998; Gao et al 1998b; Nishikawa et al 1999), which are analyzed under cyclic lateral loading through experimental or numerical techniques The ABAQUS program (1998) and a kind of beam element B21 are employed for the pushover analysis

Cantilever Box Columns with and without Longitudinal Stiffeners

Recently extensive experimental study have been carried out to survey the behavior of steel cantilever box columns with and without stiffeners, which are subjected to cyclic lateral loading as well as a constant axial load (see Fig 7(a)) A detailed summary of these studies has been reported in the literature (Usami 1996) According to this reference, the local buckling is observed to occur near the column base

in the range of about 0.7B (B is the width of the flange) or between the transverse diaphragms, if any And the mode shapes of the global and local buckling are found in the form of half sine-waves These

Ductility Issues in Thin-Walled Steel Structures

observations are in good

agreement with the effective

length assumed above and the Material E

buckling modes occurred in the Number (GPa)

analyses of stub-columns By I 197

introducing an index of ductility, II 216

8 95 / 8 y (895 is the top lateral

displacement corresponding to III 206

95% of the maximum lateral load IV 206

after peak and 8y is the yield V 206

lateral displacement), empirical

equations of ductility related to VI 206

some main parameters has been VII 206

developed for both the columns

VIII 206 with and without longitudinal

stiffeners, which are defined as

follows (Usami 1996):

T A B L E 5

MATERIAL PROPERTIES OF E X A M P L E S

v (MPa) E/gst g st /[~y

0.269 266 21.0 11.3 0.270 282 32.4 16.9

Notes: Refer to Tables 3, 4, 6, and 7 for material numbers

71

Unstiffened columns: 69s - 0.0670 + 2.60 (S = 1.09) (16)

8y [(1 + e / Py )RI ~~ ] 3"~

Stiffened columns: 89s - - = 0.0147 + 4.20 (S = 1.40) (17)

6y [(1 + P / Py )gf~~ 3"s

where S is the standard deviation; and ~ is the column slenderness ratio parameter given by

t

r ~: V E Here h is the column height and r is the radius of gyration of cross section Equations (16) and (17) were fitted corresponding to the average curve for test data (i.e., the M curve plotted in Fig 8 by the solid line) and the lower bound curves were also proposed as Eqs (16) and (17) minus the standard deviation (S), as the M-S curve shown in Fig 8 by the dashed line

Several specimens in the form of both unstiffened and stiffened columns reported in the reference (Usami 1996) are adopted here to demonstrate the validity of the ductility evaluation method proposed in this paper The parameters of the columns are presented in Tables 3, 4 and 5 The computed ductility estimations (8,,/8y) are presented in Fig 8 compared with the empirical curves (Eqs (16) and (17)), of which 89s/dy is denoted by 8u/dy for the accordance For unstiffened columns (Fig 8(a)), it is observed that the proposed method gives the ductility predictions very close to the lower bound curve (M-S

Curve), which has been recommended for the practical use considering the required safety (Interim

1996) In Fig 8(b) which is for stiffened columns, good agreement of computed results with the test curves is also observed Aswell, the previous method proposed by Usami et al (1995), where the failure strain equations based on isolated plates (Eqs (11) and (12)) are used, is also applied to these examples and the obtained results are included in Fig 8 It can been seen that the previous method underestimates the column ductility for most cases

Cantilever Columns with Pipe Sections

The behavior of thin-walled steel cantilever-typed columns with pipe section has been investigated by some researchers (e.g., Nishikawa et al 1996; Gao et al 1998b) In the cyclic test on such columns by Nishikawa et al (1996), the so-called elephant foot bulge mode was found to occur in the range of about 3.0\/R t (R is the radius of the pipe section) from the column base (Nakamura 1997) This range is

72

20

15

T Usami et al

O Previous I

I

, Cyclic Test ( U s a m i 1996) I

M curve [

i , i i i , i , i

(I+P/Py)R~ ~ (a) Unstiffened columns

.~ I O P r e v i o u s

15 i \ I ( U s a m i et al 1995)

~ ]Cyclic Test ( U s a m i 1996)

~.I0 \~, 0 l M-S

5 - ~ ~

0.] 0.2 0.3 01.4 01.5 01.6

(I+P/Py) 9 Rr" ~ o.s (b) Stiffened columns Figure 8 Ductility estimations of cantilever-typed columns with box sections

almost as same as the effective failure

length assumed in this study (Eq 15)

Through numerical cyclic analyses, some

researchers (Gao et al 1998b) proposed an

empirical equation for the ductility of

cantilever-typed columns with pipe

sections, which is given by

~9 ff_5 _ 0.24 (19)

~iy (1 + P / Py )2,3-~,3Rt

Nine such columns are investigated here,

the parameters of which are presented in

Table 6 The computed ductility

estimations are plotted in the Fig 9 by

comparison with Eq (19) It is found that

all the points corresponding to the results

of the present study lie in the vicinity of the

equation curve Thus, the applicability of

the proposed method to steel columns with

pipe sections is also verified

One-story Rigid Frame

Although the behavior of thin-walled steel

cantilever columns has been extensively

investigated by researchers, available

research findings on the thin-walled steel

frames are too limited to supply sufficient

information on the ductility evaluation

(Nishikawa et al 1999) The proposed

method is expected to be a simple but

efficient ductility evaluation tool for such

structures

A one-story rigid frame, which has been

Speci -men P1 P2 P5 P8a P8b 1897 891 P8-15 4391 891 P10 3303 580 Pll 4391 891 P12 4391 891 Notes: see Table 5 for details of material properties

TABLE 6

P a r a m e t e r s of C a n t i l e v e r Pipe C o l u m n s

(mm) (mm) (mm) Rt ~ e/ev MaterialNumber

3403 891 9.00 0.110 0.26 0.12 VI

4391 891 7.32 0.115 0.30 0.15 V

4391 891 8.41 0.100 0.30 0.15 V

2598 891 11.2 0.075 0.18 0.15 V

12.6 0.067 0.13 0.15 V 11.2 0.075 0.30 0.15 V 20.0 0.031 0.37 0.09 VII 9.61 0.088 0.30 0.15 V 16.8 0.050 0.30 0.15 V

12 \ " 9 P t I

8

4

2

(l+P/Py)Rl'S~, ~

Figure 9: Ductility estimations of cantilever-

typed columns with pipe sections

Ductility Issues in Thin-Walled Steel Structures

B

73

No VIII VIII

tested in a recent study (Nishikawa et al

1999), is analyzed through proposed

method The general layout and some

pertinent parameters of the frame are

presented in Fig 10 and Table 7 In the

beam-column connection parts, all the

panels of both beam and column

sections are strengthened by doubling

the thickness The flame was tested

under cyclic lateral loading with the

constant vertical loads of P =0.12Py at

the top of the frame

The afore-mentioned method is applied

to this structure, where it should be

noted that in a flame system, the axial

force of the columns varies with the

change of lateral load And this makes

the trial and error method required for

calculating the failure strains (see Eq

(7)) For this frame, the critical parts are

the regions marked by (~), (~), (~), @, (~)

and (~ in Fig 10 And the place where

the average compressive strain first

reaches the corresponding failure strain

is found at part (~) Figure 11 illustrates

the normalized lateral force versus

displacement curve from the pushover

analysis compared by the normalized

hysteretic curve from the cyclic test

(Nishikawa et al 1999) Both of the

points corresponding to the maximum

strength (6,.) and 95% of the maximum

strength after peak on the test envelop

curve (69s) are used to represent the

Figure 10 General layout of the frame

Figure 11 Force-displacement relation curve of the flame

ultimate state of the flame The failure points are denoted by different marks in Fig 11 It is observed that the computed ductility (ru/ry) of the proposed method is close to the 6~/~ from the cyclic test, whereas the ductility prediction based on previous method (Usami et al 1995) is too conservative In the light of safety required in practical design, the ductility capacity predicted by the proposed procedure is satisfactory

74

CONCLUSIONS

T Usami et al

The ductility of thin-walled steel stub-columns with and without longitudinal stiffeners was investigated through extensive parametric analyses The key parameters affecting the ductility of box stub-columns are found to be the flange width-thickness ratio, magnitude of the axial force, and the stiffener's slenderness ratio The effects of the cross-sectional shape and the columns aspect ratio were also investigated and found insignificant for the ductility of stub-columns On this basis, empirical equations for the ductility in terms of the failure strain were developed Besides, empirical equations of failure strains proposed for isolated plates and short cylinders in previous studies (Usami et at 1995; Gao et al 1998a) were also presented in this paper

Moreover, an evaluation procedure has been proposed to employ the ductility equations into the ductility estimation of practical steel structures composed of thin-walled box or pipe sections A simplified pushover analysis was utilized and a failure criterion was defined The procedure can be used

to evaluate the ductility of thin-walled steel structures in the form of not only cantilever-typed columns but also framing structures The proposed method was used to successfully evaluate the ductility of some cantilever-typed columns and a one-story rigid frame By the comparison with ductility estimations obtained from cyclic tests or numerical analyses reported in the literature, the reliability of the proposed method was verified

References

ABAQUS/Standard User's Manual (1998) Ver 5.7

"DIN 4114, Blatt2." (1953) Stahbau, Stabilitatsfalle (Knickung, Kippung, Beulung), Berechnungsgrundlagen, Richtlinien, Berlin, Germany (in German)

Ltd., Oxford

& Sons, Inc., New York

Gao, S B., Usami, T., and Ge, H B (1998a) "Ductility of steel short cylinders in compression and bending." a r Engrg Mech., ASCE, 124(2), 176-183

Gao, S B., Usami, T., and Ge, H B (1998b) "Ductility evaluation of steel bridge piers with pipe sections." a r Engrg Mech., ASCE, 124(3), 260-267

Nakamura, H (1997) "Formulae for evaluating shear-bending buckling strength of steel piers with circular cross section and applicability of the numerical buckling analysis method." Proc of Nonlinear Numerical Analysis and Seismic Design of Steel Bridge Piers, JSCE, 37-42 (in

Japanese)

Nishikawa, K., Yamamoto, S., Natori, T., Terao, O., Yasunami, H., and Terada, M (1996) "An

experimental study on improvement of seismic performance of existing steel bridge piers." 3' of Struct Engrg., 42A, 975-986 (in Japanese)

Nishikawa, K., Murakoshi, J., Takahashi, M., Okamoto, T., Ikeda, S., and Morishita, H (1999)

"Experimental study on strength and ductility of steel portal frame pier." a r Struct Engrg., JSCE,

45A, 235-244 (in Japanese)

Committee on New Technology for Steel Structures (CNTSS), JSCE (in Japanese)

Usami, T., Suzuki, M., Mamaghani, I H P., and Ge, H B (1995) "A proposal for check of ultimate earthquake resistance of partially concrete-filled steel bridge piers." Struct Mech./Earthquake Engrg., JSCE, 508/I-31, 69-82 (in Japanese)

HIGH-PERFORMANCE STEEL STRUCTURES:

RECENT RESEARCH

L.W Lu, R Sause and J.M Ricles Department of Civil and Environmental Engineering, Lehigh University

Bethlehem, PA 18015-3176, USA

ABSTRACT

Much effort has been devoted in the recent years to the development of high-performance structures for civil and marine construction Emphasis of this effort has been on the use of high-performance steels and innovative structural concepts to improve performance and reduce life-cycle cost The paper first gives a summary of the properties of high-performance steels available in the market This is followed by a description of research exploring application of such steels to I-girder bridges and critical elements in building structures within the framework of the current construction practice Three innovative structural concepts are then presented: a post-tensioned connection for building frames resisting seismic forces, use of high performance dampers for dynamic response control, and unidirectional double hull structure for ships Their potential applications are also discussed

KEYWORDS

High-performance structure, high-performance steel, building, bridge, ship, weldability, fracture toughness, connection, seismic resistance

INTRODUCTION

What is a high-performance structure? Presently, there is not a universally accepted answer to this question and different people are likely to provide different answers which will depend on the types

of structures the individuals having in mind, the desired levels of performance and the performance

of structures built according to the present practice No attempt, therefore, will be made to define

"high-performance." The following criteria are often used to judge the overall quality of a structure:

(1)

(2)

(3)

Performance under service load,

Performance under overload, and

Life-cycle cost

75

76 L W Lu et al

For a structure to be considered as a high-performance structure it should have one or more

improvements related to these criteria Different approaches may be adopted to achieve the desired improvements This paper is concerned with (1) the use of high-performance materials and (2) the development of innovative structural concepts to enhance overall performance and to reduce life- cycle cost

HIGH-PERFORMANCE STEELS AND THEIR PROPERTIES

A high-performance steel is defined as a steel that has the combined characteristics of high strength, good ductility, high toughness, good weldability and fabricability These are the properties essential for successful construction of high-performance structures in a civil infrastructure system For exposed structures, such as bridges and ships, good corrosion resistance is also necessary From metallurgical composition and processing point of view, a yield strength above 450 MPa is considered as high strength The fracture toughness, weldability and formability of the steels should

be significantly better than those of the conventional steels The key issue is the control of the amount of carbon and carbon equivalent (Lu, Dexter, Fisher, 1994)

The early attempts of using the traditional high strength steels in bridge and ship construction produced some unsatisfactory results These steels were found difficult to fabricate due primarily to susceptibility to hydrogen cracking and the risk of brittle fracture associated with materials having inadequate fracture toughness Other problems include: 1) welding defects other than hydrogen cracking, and 2) potential for stress-corrosion cracking Many bridges fabricated in the 1960's and early 1970's from ASTM A514/A517 (690 MPa yield) steel suffered from hydrogen cracks which occurred during fabrication (Fisher, 1984) Many of these hydrogen cracks occurred in the longitudinal web/flange joint of welded built-up box sections used as tie girders in tied arch bridges (Anon, 1979, Fisher, Pense and Hausammann, 1982) as well as welded built-up plate girders One example is the Gulf Outlet Bridge near New Orleans Some bridges have also experienced hydrogen cracking in transverse groove welds, e.g the 1-24 bridge over the Ohio River near Paducah, Kentucky (Fisher, 1984) Hydrogen cracking was also observed in the Navy's Seawolf submarine in the 120-S weld metal used with the 690 MPa yield strength HY-100 steel (Anon, 1991)

Hydrogen cracking is most effectively avoided by using steel and weld metal with microstructures that are not susceptible It has been shown that susceptibility to hydrogen cracking increases significantly as the carbon content exceeds 0.1 percent (Graville, 1976) The susceptible microstructures are typically martensite The new high-performance steels with low carbon contents are not susceptible to hydrogen cracking

Microalloyed steels with low carbon content, high manganese levels and microalloy carbide and nitride formers have been available for sometime for use in construction of structures that require high strength, high fracture toughness, and good weldability Over the past 15 years, low-carbon, age-hardenable steels have gained increasing usage in shipbuilding, heavy-vehicle manufacturing, and offshore structure construction because of their excellent weldability and fracture toughness These steels have become known as High-Strength Low Alloy (HSLA) steels although their total alloy content is generally around four percent Another method of increasing strength without increasing carbon and alloy content is controlled rolling combined with on-line accelerated cooling, i.e thermo-mechanical controlled processing (TMCP)

These high-performance steels offer some clear benefits when compared with the traditional high strength steels (Bolliger, et al, 1988) Most are virtually immune to hydrogen cracking in the heat- affected zone (HAZ) of welds This superior resistance to hydrogen cracking allows these steels to

High-Performance Steel Structures: Recent Research 77

be welded without the application of preheat in most situations The low-carbon, fine-grained microstructure that results from typical processing yields a very favorable combination of high strength and high toughness The excellent fabricability, strength and toughness make high- performance steel very attractive for use in many applications For bridges, these advantages may allow consideration of lifting the onerous requirements for fabrication of fracture critical members (FCM) FCM are members subjected to tension which if fractured will cause failure of the structure The following are some examples of the currently produced high-performance steels:

Low Carbon Age-Hardening Nickel-Copper-Chromium-Molybdenum-Columbian and Nickel-Copper-Columbian Alloy Steels, ASTM A710

High Yield Strength, Age-Hardening Alloy, Structural Steels (HSLA 80 and HSLA 100), MIL-S-24645A

Structural Steel for Bridges, ASTM A709 Grade HPS 485W

There are several copper-nickel high-performance steels for bridge construction under development

at the ATLSS Center of Lehigh University (Gross, Stout, and Dawson, 1998)

APPLICATION OF HIGH-PERFORMANCE STEELS

A substantial number of studies have been carried out in the ATLSS Center and elsewhere to explore the use of high-performance steels in bridges, buildings, offshore structures, and ships Brief descriptions of three of these studies are given below:

1-Girder Highway Bridges

The advantages of using high-performance steel in conventional I-girder highway bridges has been investigated by Sause and Fisher (1995) The investigation involved redesign of recently constructed highway bridges, using conventional steels with yield strengths of 250 MPa and 345 MPa, and using high-performance steels with yield strengths between 485 MPa and 825 MPa The normalized weight of minimum weight girder cross-sections designed for each steel is plotted versus yield strength in Figure 1 The weight of the design using 345 MPa steel is taken as 100%, and the weight of the designs using other steels are normalized by the weight of the 345 MPa design Three cases are considered: (1) design for strength and stability according to the AASHTO specifications without considering fatigue, indicated by the dashed line with circles; (2) design for strength and stability without considering fatigue and allowing the plastic moment to be used as the nominal bending strength of compact girder cross-sections, indicated by the dashed line with squares; and (3) design considering strength, stability, and fatigue indicated by the solid line with solid boxes As seen in Figure 1, if fatigue is not considered, a higher steel yield strength usually results in a smaller weight per length However, an exception occurs at 485 MPa because the AASHTO specifications permit the use of the plastic moment as the nominal bending strength of compact girder cross-sections only when the yield strength is no more than 485 MPa As a result, higher strength steel girder cross-sections must be designed with the yield moment (yield stress at the extreme fiber) as the nominal bending resistance This limitation in the design specifications is based on concern about the ductility of structural members fabricated from high-strength steel Girders fabricated from high-performance steel may not require this limitation, although further study of this issue is needed The dashed curve with squares in Figure 1 represents the case when the plastic moment is used as the nominal bending strength of all compact cross-sections The solid

78

120

L IV Lu et al

1 1 t I i

lOO .s

-~ 80

Figure 1 Weight versus yield strength for minimum weight steel I-girder cross-sections line shows that when fatigue of welds between transverse stiffeners and the web and flange plates is considered in design, potential decreases in weight with increasing yield strength end at a yield strength of 690 MPa, because of stress range limits for the details

In addition to stability and fatigue, deflection under live load may also be a design constraint The elastic live load deflection of I-girder bridges designed using high-performance steel was considered by Sause and Fisher (1995) AASHTO deflection criteria were applied to high- performance steel bridge designs to investigate whether these criteria are constraints on the use of high-performance steel Live load deflections were calculated for minimum weight bridge designs developed for each yield strength level, and plotted versus yield strength level in Figure 2 The deflection limit is L/800 Figure 2 shows that the bridge designs at each strength level satisfy the deflection limit However, the assumptions made in computing the live load deflections according

to AASHTO may not be acceptable to many bridge engineers With more conservative assumptions, the computed deflections for bridge designs at the highest yield strength levels may exceed the deflection limit

E

E 50

C

0

25

9 Design for Mp [] Design for My

D e f l e c t i o n Limit

0 I I ! I I

(275) (415) (550) (690) (830)

2 E

r-

.s

Vleid Stress, ksi (MPa)

Figure 2 Live load deflection versus yield strength for minimum weight bridge designs

Connections in Building Frames

The superior ductibility, toughness, and weldability of the high-performance steels make them ideal

High-Performance Steel Structures: Recent Research 79 material for critical elements in structural systems Examples of critical elements, where such properties are required, include connecting plates in beam-to-column connections, connectors or connecting devices, shear links in eccentrically braced frames, tension members of structures in severe service environments, etc These steels are also attractive for large and complex structures because of the possibility of requiring no pre- and post-weld treatment

A study of the use of A710 steel plates as the flange connecting plates in a beam-to-column web connection shown in Figure 3 was carried out An identical connection, but with ASTM A572 (50) steel plates, was tested at Lehigh University It failed prematurely due to fracture of one of the connecting plates The fracture was predominately brittle in nature although there was evidence of several crack arrests which indicate some ductility in the region The factors contributing to the fracture include: (1) a large amount of plastic strain imposed on the members, (2) strain concentrations at design details, and (3) orientation of the plates with the applied strain in the least fracture-resistant direction (the rolling direction was parallel to the fracture plane) A post-test examination showed that the defects in the weldments were no greater in size than might be found

in typical structural welds It is felt that this connection fractured in a brittle manner due to the large applied tensile strain which was concentrated at the design detail

Figure 3 Beam-to-column connection test details with A710 steel

For the A710 tests, the flange plates was orientated so that the applied strain was not in the direction

of the least fracture resistant direction; the rolling direction of the plate was parallel to the applied tensile strain The strain concentrations at the design details were difficult, if not impossible, to avoid in construction The connection was, therefore, assembled as if in an actual construction environment This specimen behaved in a very ductile manner and the ultimate load exceeded the calculated plastic limit load by about 20% (Lu and Fisher, 1990)

The ATLSS Center has developed a wedge and socket type joint for a beam-column connection in a