Effects of tunnel construction on nearby pile foundations 2

1.1.1 Tunnel construction near pile foundation Among all the passive loading caused by soil movement, the one caused by tunnelling is perhaps by far the most complicated.. In order for t

The increased demand for underground systems in urban areas particularly for mass transportation can be seen in big cities like Singapore, Hong Kong and London This has led to many tunnels being constructed in proximity to structures Extensive research has been carried out in the United Kingdom particularly on the Jubilee Line Extension (Burland et al., 2002) on the effects of tunnelling on nearby structures However, most of the structures in the studies are supported on shallow foundations and very little work has been carried out on structures supported on pile foundations This is due to the fact that most structures were built long before the tunnels are planned As a result, instrumentation inside existing pile foundation is difficult to install for further investigation

1.1.1 Tunnel construction near pile foundation

Among all the passive loading caused by soil movement, the one caused by tunnelling is perhaps

by far the most complicated This is due to the complex tunnelling processes particularly for shield tunnelling, which comprises shield machine advancement, application of face pressure, tail void grouting and lining installation These caused significant disturbance to the surrounding soil with shearing (during shield advancement), loading (application of face pressure and grout pressure) and unloading (soil stress relief) mechanism Since the structures usually exist in an urban environment long before a tunnel is planned, engineers only have the choice of aligning tunnel position relative to the nearby pile foundation Figure 1.1 shows a schematic illustration of two typical situations being encountered in practice: (a) tunnelling under pile foundation and (b) tunnelling adjacent to pile foundation

During tunnelling, stress relief will occur in the surrounding soil When the tunnel is constructed under a pile, it is likely that the pile base resistance will be first reduced and in turn leads to pile settlement To maintain equilibrium of load, the base resistance is transferred to the pile shaft If the tunnel stress relief is great enough to fully mobilise the pile shaft resistance, a larger pile settlement would be anticipated However, pile lateral response is unlikely to be of significance in this situation

In the second situation where the tunnel is constructed adjacent to a pile foundation, a different mechanism is observed The stress relief due to tunnelling would cause soil settlement above tunnel This in turn causes negative skin friction (NSF) to act along the pile shaft above the tunnel level In order for the force to be in equilibrium, the pile shaft below tunnel level (which is not subjected to settlement) would support the dragload from NSF above the tunnel level Only when the positive shaft resistance and pile base are fully mobilised, settlement would become a problem

This depends on the availability of pile length extension below tunnel level It is also to be checked that the dragload would not cause the pile to overstress In addition, the lateral pile response can be significant since horizontal soil movement is largest near tunnel

Besides the two possible situations, other possibilities of tunnel alignments near pile foundation were identified from reported case histories and can be grouped into four different categories as shown in Figure 1.2 The categories are based on the relative position of tunnel to pile foundation The pile responses due to tunnelling in each category requires further investigation owing to the varying observations by previous researchers as presented in Chapter 2

The ignorance of not taking into account the additional loading caused by soil movement as mentioned above could lead to excessive pile settlement or pile structural capacity being exceeded Under-designed pile foundations will be reflected on the superstructure such as cracks on beam, column or wall and ultimately collapse if the damage is large Aside from potential damage, poor understanding of the mechanism will also lead to an expensive protective or mitigation work Therefore, further studies would be required to develop a better understanding of the problem and contribute to an economic design

1.1.2 Current design and construction approach

The current codes of practice do not provide guidance and basis for the design of piles subjected to soil movement caused by tunnelling The design requirements are usually stipulated by the local authorities To date, there are a few design approaches available to analyse the problem Figure 1.3 summarises the approaches obtained from literature review In the direct method, pile responses are computed from some analytical and numerical analyses These include finite element analysis, boundary element analysis and soil-spring analysis Besides, design charts are also available In

the indirect method, pile responses are not computed For example, the ‘risk of damage to building’ assessment method categorises the overall building damage, whereas methods used by Jacobsz et al (2005) assumed the pile to response similarly to greenfield soil movement Furthermore, Nakajima et al (1992) and Inose et al (1992) check the factor of safety of pile bearing capacity by hand calculation More details of the methods mentioned are described in Chapter 2

Despite the design and assessment carried out, some local authorities also impose restrictions on the tunnelling activities near critical structures In Singapore, the Land Transport Authority (LTA) imposes a criterion for the design of deep foundation to allow for an additional movement of 15mm in the short term and 25mm in the long term to account for future development (LTA, 2002) In addition, LTA defines an area of first reserve which is typically 6m away from the tunnel extrados Within this area, the activities should be dealt with care In Japan, the provision of tunnelling work adjacent to foundation can be classified into three zones (Figure 1.4) as according

to Fujita (1989) If the tunnel lies within Zone 1, there is no work restriction In Zone 2, the tunnelling work shall proceed with prudently selected methods and techniques In Zone 3, auxiliary measures and prediction should be carried out Similar classification was also adopted by Moroto et al (1995)

1.2 Objectives of the study

With the current understanding of the effects of tunnelling near pile foundation based on the limited available case histories, laboratory tests and numerical studies, the design of tunnel near pile foundation is yet to be understood Despite the increasing demand for underground systems, the constraint in the congested area inevitably creates pressure for tunnelling engineers to make

sure that the construction of tunnel does not cause detrimental effects to adjacent pile foundations The thought and potentially wide scope have naturally stimulated the eagerness to undertake the current study The aim of the thesis is to develop a better understanding of the effects of tunnel construction on nearby piles and hence contribute to the design and analysis of such piles This aim is achieved through the study using various methods as outlined below

As mentioned above, placing instrumentation inside existing pile foundation is not feasible, which restricts further understanding of pile responses caused by tunnelling Therefore, it is one of the objectives of this research to carry out field study with in-pile instrumentation A Mass Rapid Transit (MRT) line constructed in Singapore, North-East Line (NEL) C704 was identified for this study In the NEL C704, twelve working piles were instrumented with extensive strain gauges which allow the bending moment and axial force developed during tunnelling to be monitored The case study will add to the existing knowledge of field monitoring of pile responses during actual tunnelling

The case study was only limited to specific range of tunnel-pile configurations, tunnelling process and soil type Further understanding of the problem outside the range is therefore unknown To overcome the limitation, three-dimensional (3-D) finite element (FE) studies were performed The

FE model allows varying parameters and configurations to be studied In order to do a general study, a reliable finite element model has to be established first Due to the extensive monitoring data of the NEL C704, a 3-D finite element model was set-up to back-analyse the case study The aim of the back-analysis is also to verify the reliability of the field monitoring data Subsequently,

a simplified 3-D model was used for parametric studies to identify the critical and non-critical aspects in the analysis of pile responses due to tunnelling

Although 3-D finite element model is a powerful tool for analysing such a complex interaction problem, 3-D software is usually not adopted in practice for design due to the limitation of computational resources available and also the time consuming effort Therefore, the problem is usually idealised in plane strain condition in 2-D analysis The plane strain idealisation is just an approximation to the actual 3-D nature of the problem The idealisation techniques available are not well understood and could be misused since it is difficult to select adequate parameters to represent the 3-D effect Besides, there are very few studies being carried out on 2-D idealisation

of tunnel-pile interaction Therefore, the current research would also provide greater insight into the use of 2-D finite element model in analysing the problem

1.3 Organisation of thesis

Firstly, a literature review on the current research topic is presented in Chapter 2 The review consists of past observations from case histories, laboratory tests and field studies This is followed by a review on the various prediction and design methods available in practice The current understanding and outstanding issues were also discussed

Chapter 3 presents a unique case history in Singapore on the monitoring of the effects of tunnel construction adjacent to full-scale working piles The field instrumentation data were studied and presented together with the project overview, ground condition and construction details Also presented is the initial analysis carried out using existing design charts

Chapter 4 presents the development of a three-dimensional finite element model to back-analyse the case history reported in Chapter 3 Full tunnel construction process was simulated The analysis results for single tunnel advancement, twin tunnel advancement, pile response to single

tunnel and twin tunnels were discussed Sensitivity studies were also carried out, investigating various factors affecting the 3-D model results

Chapter 5 gives a general overview on some of the considerations required in finite element analysis of piles due to tunnelling Parametric studies were carried out to cover a wider range of tunnel-pile configurations beyond the case study examined in Chapter 4

Chapter 6 presents the study of tunnel-pile interaction problem using plane strain idealisation technique Some of the techniques available for modelling pile foundation and tunnelling were reviewed followed by thorough comparative study between 2-D and 3-D models Sensitivity studies were carried out and calibration charts were produced to serve as a guide for analysing the tunnel-pile problem in plane strain model Three case studies were analysed using 2-D FE model, engaging the technique and charts presented earlier

Chapter 7 lists the summary of findings and conclusions from the research work carried out in this

thesis Recommendations for future research areas are identified

Section 2.3 discusses the design and prediction methods that have been used by engineers to analyse the problem The advantages and shortcomings of each method are highlighted Based on the limited publications available, the current understanding and outstanding issues are identified and discussed in Section 2.4

2.2 Pile responses caused by tunnelling: Physical observations

Engineers have reported evidence of case histories where tunnels were constructed close to pile foundation Table 2.1 summarises some of the case histories and its details including the distance between tunnel axis and pile centre (Xpile), tunnel depth (Htun), tunnel diameter (Dtun), pile diameter (Dpile), pile length (Lp) and volume loss (VL)

In Singapore, construction of the MRT North-East Line (NEL) C704 was a unique case for studying the responses of pile subjected to tunnelling (Coutts & Wang, 2000) As part of the contract C704, a viaduct bridge was planned in conjunction with the tunnel advancement The bridge which consists of 2 abutments and 39 piers was constructed in parallel alignment with the new twin tunnels configuration (Figure 2.1a) The piers were supported by groups of four to six 1.2m and 1.8m diameter bored piles Along the alignment, 6.5m diameter tunnels were located very closely with 1.6m clear distance to the pile foundation The piles were founded at relatively greater depth to tunnel level with Lp/Htun ratio of 3 (i.e long pile) A total of twelve piles were installed with strain gauges at various levels where the axial force and bending moment were obtained Only a summary of the monitoring results for six piles were reported Substantial dragload and bending moment were observed in the piles, as high as 91% and 59% of the design working load and bending moment respectively Unfortunately, many important details such as tunnel volume loss, pile-tunnel configurations, construction sequence and soil data were not mentioned by the authors, which lead to limited interpretation Despite that, further collection of data from the contractor and the authority was carried out as part of the works in this thesis and formed the study as presented in Chapter 3

Tham & Deutscher (2000) reported another stretch of the MRT NEL tunnel (Contract C705) passing by a 4-storey workers’ quarters supported on 0.45m diameter bored piles with tunnel-pile clear spacing of approximately 1.85m (Figure 2.1b) The tunnel was located at a depth of 14m.b.g.l which correspond to the same length of the piles (i.e Lp/Htun of 1) Comprehensive monitoring by building settlement markers showed that the building did not suffer detrimental effect Settlement of up to 7mm was measured at the edge of the building The success of the operation was due to the good tunnelling control where volume loss was only 0.4%

In Hong Kong, construction of the 7.9m diameter twin tunnels for Mass Transit Railway (MTR) Island Line posed the same concern to adjacent piled-building which was located approximately 3m to the tunnel extrados (Forth & Thorley, 1996) The 31-storey building was supported on 2m diameter bored pile with pile group consisting of up to 4 piles The Lp/Htun ratio was between 1.6 and 2.5 which were considered as long pile (Figure 2.2) Again, the performance of piles could only be judged indirectly from building movement at ground surface Settlement of only 5mm was observed on the near side of the building

In London, Mair (1993) and Lee et al (1994) reported the construction of a hand-dug escalator tunnel at the Angel Underground Station where the tunnel was constructed very close to pile foundation (i.e 1m clear distance between tunnel and pile) The new 7-storey building was supported by 1.2m diameter under-reamed bored piles (Figure 2.3) which were installed below the tunnel depth hence, a long pile condition These piles were de-bonded down to 4m above the pile tip to reduce negative skin friction above the tunnel level Owing to the early planning and co-operation between tunnel owner and building developer, both in-ground and in-pile instrumentations were possible prior to tunnelling Measurements of the in-pile inclinometers showed that the nearest pile was only subjected to maximum lateral deflection of 8mm for volume loss up to 2% Besides, both the in-ground and in-pile inclinometers results were very similar The

authors concluded that tunnel could be constructed very close to pile foundations in London Clay and would only cause small horizontal deflection

Powderham et al (1999) reported construction for the Jubilee Line Extension (JLE) station beneath piled-supported buildings Figure 2.4a shows sectional view of the tunnels and buildings Generally, the tunnels were driven below the 20m long bored piles Prior to tunnelling, permeation grouting was carried out around the tunnels and one of the buildings was underpinned with concrete raft During tunnelling, compensation grouting was applied to minimise settlement Maximum settlement of only 18mm was observed after the two tunnels were driven The risk management approach used was able to limit the damage of buildings to acceptable levels For the same tunnels, Selemetas et al (2002) reported the settlement of New London Bridge House (NLBH) which is adjacent to the buildings mentioned by Powderham et al (!999) The 1.4m diameter under-reamed pile supporting a corner of the NLBH was located 2m above the tunnels but not directly under the NLBH (Figure 2.4b) Monitoring of NLBH showed a settlement of up to 24mm after the tunnels were constructed The authors concluded that no adverse effects were encountered on the piled-structure

Recent tunnelling works for the CTRL2 in London also exposed piled-structures to potential damage Jacobsz et al (2005) reported the monitoring of three piled-bridge foundations with one

on end bearing piles (Figure 2.5a) and two on friction piles (Figures 2.5b and c) The authors recommended that re-assessment of pile capacity to be carried out as large factor of safety can often be found in piles and redistribution of loading is possible

In Japan, construction of a tunnel near a highway bridge posed the same concern (Moroto et al., 1995) Piles supporting the bridge piers are 15.5m away from the tunnel axis (Figure 2.6) No settlement or tilt was observed at the piers while maximum surface settlement of 6mm was

measured For the construction of Nanboku Line in Tokyo, Nakajima et al (1992) and Inose et al (1992) described the planning of tunnels below piled bridges (Figure 2.7) and a large piles supported dome stadium (Figure 2.8) respectively Pile settlement and bearing capacity were of major concern during tunnel construction due to its relative position directly below tunnels In similar condition, Ikeda et al (1996) made use of compensation grouting between tunnels and pile foundations to restrict settlement of piles More recently, Takahashi et al (2004) reported the construction of Rinkai Line in Tokyo where twin tunnels were driven underneath pile foundations with minimum clearance of 3.4m Figure 2.9 shows the relative location of tunnels and pile foundations With the use of mitigation measure such as grout injection and good control of tunnel construction which limit the volume loss to 0.5%, settlement of bridge pier supported by the piles was monitored to be less than 4mm

2.2.2 Laboratory and centrifuge tests

As described in Section 2.2.1, only two out of the many case histories were installed with in-pile instrumentation The pile responses from the remaining case histories could only be examined indirectly from the monitoring of the structures Therefore, some researchers resorted to experimental simulation where pile can be instrumented to obtain valuable data

One of the earliest studies on pile responses due to tunnelling was reported by Morton & King (1979) In their study, four 1-g static model tests were carried out in medium dense to dense dry sand The tunnel was excavated with jacking liner tube by concurrently rotating a close fitting full-face head Wooden piles with sustained working load were located at three levels above the tunnel crown (Figure 2.10) in separate tests Despite the violation of scaling law, the authors have concluded that the settlement of friction pile could be large as the pile tip is closer to the tunnel Besides, the prime factor of pile failure is due to dilatancy of soil within a zone above the tunnel

The model indicated that even small dilatancy can induced immediate failure of the piles located within the critical zone However, the results were questionable due to the 1-g representation of real problem where the low confining stress in sand (which is stress-dependent) caused an abrupt failure in pile without gradual increase in settlement (Mair, 1979)

With the availability of centrifuge modelling technique, a prototype problem can be simulated in a scaled laboratory model hence overcoming the limitation of 1-g model Some of these centrifuge tests on tunnel-pile interaction have been carried out in clay overlying dense sand (Bezuijen & Schrier, 1994; Hergarden et al., 1996), stiff clay (Loganathan, 1999), soft clay (Ran et al., 2003), dense dry sand (Jacobsz et al., 2002; Feng et al., 2002) and dense saturated sand (Lee & Chiang, 2004) Table 2.2 summarises all the centrifuge tests and its details It should be noted that all the tests considered a plane strain tunnel in the simulation Besides, most of the tests were simulating volume loss as small as 1% and up to 20% to investigate the pile responses over more detrimental tunnelling situation

The three tests carried out by Bezuijen & Schrier (1994) (see also Hergarden et al., 1996)

simulated a prototype tunnel of 7m diameter at a depth of 14.5m, 18m and 23m below ground surface Six end-bearing piles of 0.4m diameter were installed to a depth of 18m below ground surface (Figure 2.11) The six piles corresponded to four different tunnel-pile distance (Xpile) of 4.9m, 6.5m, 9.7m and 12.9m The authors concluded that pile settlement can be significant when volume loss is 1% or more and the clear distance between tunnel and pile is less than one tunnel diameter Furthermore, settlement and bearing capacity of pile is not affected at all when the clear distance is more than two times the tunnel diameter

Loganathan (1999) carried out three centrifuge tests which have quite similar tunnel-pile relative vertical position except that the clear distance between tunnel and pile was fixed at 2.1m Besides,

both single pile and 2x2 pile group were investigated in stiff clay (see Figure F.1) In the tests,

both the axial and bending responses of the piles were measured The author concluded that bending moment (BM) and lateral deflection of a pile is critical when tunnel springline is located

at or near the pile base The axial force is found to be critical when the tunnel springline is below the pile base Besides, the BM and lateral deflection of both single pile and pile in the group at the same distance from tunnel are almost identical

Studies by Jacobsz et al (2002) focused on the investigation of settlement and load distribution of single driven pile in homogeneous dense sand All the tests were carried out with the pile base above the tunnel The zone of influence to determine the pile’s critical position was identified as shown in Figure 2.12 (shaded area) As volume loss increases, load transfer from pile base to shaft changes gradually and once full mobilisation of shaft resistance has been achieved, large settlement occurs Pile base located within the zone would be subjected to large settlement if volume loss increases beyond 1.5%

Lee & Chiang (2004) also simulated the problem in sand but concentrated on pile base located outside the zone of influence A total of twelve tests were carried out on single piles with varying pile length and tunnel depth From the observation of unit skin friction development during volume loss, load transfer mechanism was identified for long pile (i.e Lp/Htun > 1.0) and mid-length pile (i.e Lp/Htun = 1.0) as shown in Figure 2.13

Preliminary centrifuge tests were also carried out in National University of Singapore by Feng et

al (2002) and Ran et al (2003) However, the test set up uncovered a problem in simulating actual tunnel stress relief Instead, an ovalisation shape of tunnel deformation was observed which caused unusual soil movement around tunnel and could not be verified further

2.2.3 Full scale pile tests

Despite the usefulness of laboratory tests as mentioned above, the simplification (i.e plane strain tunnel) and boundary problems (i.e boundary effect and drainage) of such tests raised uncertainties in the actual pile responses

In view of the construction for North/South Line in Amsterdam where tunnels have to be driven under more than 250,000 wooden and 2000 concrete piles, full scale pilot test was first carried out

A total of 43 timber piles and 20 concrete piles were purposely installed and loaded at the Second Heinenoord tunnel site (Teunissen & Hutteman, 1998; Kaalberg et al., 2005) Figure 2.14a shows the piles and tunnel layout Pile and soil settlement were measured and the results were concluded

in the zone of influence as shown in Figure 2.14b Basically, pile settlement would be greater than soil surface settlement if the pile base is founded in Zone A If the pile base is in Zone B, both the pile and soil surface settlement would be equal Where as if it is in Zone C, pile settlement would

be significantly lesser than the soil surface settlement

Selemetas et al (2005) reported another full scale trial pile test for CTRL Contract 250 in the U.K Four purposely installed piles (2 end bearing and 2 friction piles) were instrumented and loaded with kentledge These 0.48m diameter piles of 13m length were located strategically to investigate the zone of influence as categorised by Jacobsz et al (2002) Figure 2.15 shows the categorisation

of influence zone from the field data The categorisation is slightly different from Jacobsz et al (2002) and Kaalberg et al (2005) mainly by the angle of zone

2.3 Pile responses caused by tunnelling: Prediction and design methods

Given the complexity of pile responses due to tunnelling, prediction and design would not be straightforward For simplicity, empirical methods are adopted (Tham & Deutscher, 2000; Nakajima et al., 1992; Jacobsz et al., 2005) However in recent years, various methods have been used to analyse the problem such as the 2-D finite element method (Vermeer & Bonnier, 1991; Lee et al., 1994), 3-D finite element method (Mroueh & Shahrour, 2002; Cheng et al., 2004; Lee

& Ng, 2005), numerical & analytical methods (Broms & Pandey, 1987; Chen et al., 1999; Sawatparnich & Kulhawy, 2004; Kitiyodom et al., 2004) and design charts (Chen et al, 1999) Table 2.3 summarises some of the reported studies and their details

Jacobsz et al (2005) described the empirical methods used to assess pile foundation supporting three bridges (Figure 2.5) near the CTRL2 tunnels The methods used to assess pile settlement and pile overstress are summarised in Figures 2.16a and b respectively A few assumptions were made

in the assessment based on observations from centrifuge tests by Jacobsz et al (2005) and field

study by Selemetas et al (2005) The method is applicable to pile with its base located in the zone

of influence Besides, only pile axial response can be assessed

Nakajima et al (1992) reported some of the methods used to assess pile bearing capacity during shield advancement and due to tail void grouting The procedure is summarised in Figures 2.16c and d respectively Besides, Inose et al (1992) adopted an alternative approach to assess the pile bearing capacity based on an imaginative cone around the pile base (see Figure 2.16e) These methods can only be used for pile base located above tunnel

2.3.2 Finite element method

The empirical methods as described above can only be used to assess pile response individually There is a need for unified approach to analyse both the pile axial and lateral responses simultaneously such as finite element (FE) method

Vermeer & Bonnier (1991) carried out 2-D FE analyses using PLAXIS to simulate a typical pile settlement problem due to tunnelling in Amsterdam A row of piles were modelled as sheet pile wall using line elements whereas the pile-soil interface was modelled with joint elements The method used to reduce the pile stiffness and skin friction was described However, the accuracy of the method is unknown and not investigated It was concluded that the pile settlement followed exactly the settlement of bearing layer where the pile base were founded

Lee et al (1994) made a prediction for the Angel Underground Development as described in Section 2.2.1 using 2-D FE program “OASYS SAFE” Linear elastic soil model and undrained condition were assumed The piles subjected to tunnelling were not modelled directly, but were

assumed to act as slender members that deform with soil The authors claimed that the FE prediction of pile lateral deflection is similar to the measured and provided an upper bound value

Owing to the limitations of 2-D analysis as described, 3-D FE analyses were performed Mroueh

& Shahrour (2002) carried out simulation of 3-D shield tunnel advancement on adjacent pile foundation Instead of the full procedure for shield advancement, a simplified stress release zone was simulated to represent the effect due to shield, over-cut and tail void grouting No physical data was back-analysed and a reference case was set up for parametric studies The results show that tunnelling induced significant deflection and axial force in piles depending on the location of pile base Positive effect with reduction of axial force in rear piles was observed However, there was no significant reduction in bending moment Besides, the presence of pile cap only affects the upper portion of the piles and analysis can be performed assuming free head for pile group

Lee & Ng (2005) also carried out a 3-D FE analysis where the responses of single pile due to an open face tunnel (i.e no face pressure and unsupported length of 3m in front of tunnel) was studied The model was developed to mimic one of the centrifuge tests by Loganathan (1999) where the geometry and tunnel-pile configuration were same However, the ground was modelled

as London Clay which does not give any basis for comparison Besides, the simplified model does not represent a shield tunnel nor a typical NATM tunnel and certainly not a plane strain tunnel as modelled in the actual centrifuge test Despite that, some importance observations can be drawn from the analysis Firstly, the pile factor of safety was reduced from 3.0 to 1.5 after the tunnel had advanced past the pile (Figure 2.17) Secondly, a significant zone of one tunnel diameter ahead and behind the tunnel was identified where pile settlement was greater than soil surface settlement Thirdly, the transverse bending moment was three times larger than the longitudinal bending moment Finally, the induced axial force and bending moment were not significant

Cheng et al (2004) developed a Displacement Control Method to simulate tunnelling adjacent to

single pile in 3-D FE model In the model, plane strain tunnel was simulated instead of tunnel advancement Parametric studies were carried out and revealed that bending moment is negligible when Xpile/Dtun > 2 Furthermore, pile cracked moment is exceeded when Xpile/Dtun < 1 The axial force was found to be dependent on pile base position, soil stiffness and volume loss The field data from Coutts & Wang (2000) was back-analysed and good agreement of bending moment and

axial force between measurement and FE analysis was obtained

2.3.3 Numerical and analytical methods

Broms & Pandey (1987) reported one of the earliest numerical studies dedicated to pile response due to tunnelling A 2-D FE model was adopted Generally, the FE model was first used to compute the greenfield lateral soil movement at the pile position Subsequently, the lateral soil deflection was converted to an equivalent lateral load and was imposed on the beam on elastic foundation The beam was assumed to be supported on a series of Winkler springs In the approach, the pile was considered to be infinitely stiff and pile-soil-tunnel interaction was not taken into account A series of design charts were prepared for the assessment of pile bending moment Maximum bending moment was observed near pile cap whereas the bending moment was small at tunnel springline level

Chen et al (1999) also carried out a similar analysis as Broms & Pandey (1987) The lateral and axial responses of a single pile were analysed in two-stage approach In the first stage, vertical and lateral greenfield soil movements were computed from an analytical method In the second stage, the computed soil movements were imposed on boundary element analyses (i.e PIES and PALLAS) to compute the pile responses The authors noted that the lateral and axial pile

responses computed separately would lead to a lower predicted bending moment Parametric studies were also carried out and found that pile responses are influenced by tunnel-pile geometry, volume loss, soil stiffness and strength The same study was extended to investigate the influence

of pile head condition in Chen & Poulos (1997) Pile head condition has a significant effect on bending moment in pile especially when the pile head is fixed from translation and restrained from rotation Besides, a rigorous 3-D boundary element program GEPAN was further developed to couple the pile lateral and axial responses The above studies were then extended to pile group and was reported in Loganathan et al (2001)

Surjadinata et al (2005) continued to adopt the exactly same procedure as described by Chen et al (1999) However, in the current two-stage approach, 3-D FE analysis was used to compute the greenfield soil movements instead of using analytical method

Kitiyodom et al (2004) developed their own numerical program called PRAB to model the effect

of tunnelling on single pile, pile group and pile raft The approach used was similar to Loganathan

et al (2001), i.e 2-stage approach In the program, piles, soil and raft were modelled as elastic beam, springs and thin plate respectively The program was verified by comparison of results from GEPAN (Loganathan et al., 2001) and FLAC (Matsumoto et al., 2005) Parametric studies were also carried out and the authors concluded that single pile analysis can be used to represent the results of piles in pile group (i.e bending moment, lateral deflection and settlement) except pile axial force The author also confirmed that this is only true for large pile slenderness ratio

Lastly, Sawatparnich & Kulhawy (2004) presented the similar two-stage approach used by Broms

& Pandey (1987) The greenfield soil lateral deflection was first computed using analytical method

by Loganathan & Poulos (1998) Then the soil lateral deflection was imposed on the pile modelled

as beam element and soil-pile interaction modelled as springs However, a more complex

hyperbolic force-displacement curve was developed to represent the springs The authors carried out a set of parametric studies on pile bending moments and concluded that Lp/Htun and tunnel radius have great influence on the results

From the two-stage approach mentioned above, Chen et al (1999) carried out a thorough parametric study and developed a set of design charts for use to estimate the maximum pile responses due to tunnelling such as the induced bending moment, axial forces, settlement and lateral deflection Corrections for soil undrained shear strength, pile diameter, pile length to tunnel depth ratio were considered

2.4 Current understanding and outstanding issues

Research works carried out to date had shown some similarities and variations in their results The following sections present consecutively the comparison of all the reported field observations, results of pile settlement, axial force, lateral deflection, bending moment and pile group effect Outstanding issues were identified and discussed

2.4.1 General

The case histories with various tunnel-pile configurations as shown in Section 2.2.1 have indicated all possible situations encountered in practice As noted, only two case histories were reported with in-pile instrumentation, i.e Mair (1993) and Coutts & Wang (2000) These data are

invaluable and serve as references to others All the cases showed some common observations as follow:-

• Tunnelling with well controlled volume loss (typically 0.5% to 1%) caused no significant effect on the piled-structure particularly the settlement

• In some cases, measures such as compensation grouting (Ikeda et al., 1996; Takahashi et al., 2004), mass concrete underpinning (Powderham et al., 1999), pile shaft grouting (Jacobsz et al., 2005) and slip coating on pile (Mair, 1993) were adopted to minimise the effects on piles or directly on structures

• Other similar cases reported by Coutts & Wang (2000), Tham & Deutscher (2000) and Forth & Thorley (1996) showed that piles were able to withstand the tunnelling induced movement without any mitigation measure and no detrimental effect was observed

The varying opinions on similar problems have raised questions on the confidence of tunnelling near to pile foundations and the necessity for expensive mitigation works More field monitoring data are therefore required

Centrifuge tests by Loganathan (1999) showed that the largest pile settlement occurs when Lp/Htun

is equal to 1.0 and smallest when Lp/Htun is greater than 1.0 Interestingly, settlement for Lp/Htun < 1.0 is less than settlement for Lp/Htun = 1.0 (Figure 2.18a) Centrifuge tests by Bezuijen & Shrier (1994) and Hergarden et al (1996) also observed similar trend However, in Hergarden et al (1996), the trend reversed when Xpile/Dtun is increased This is believed to be due to the change of pile base position relative to the zone of influence Besides, it is found that pile settlement is significant if the pile-tunnel distance normalised with the tunnel diameter, Xpile/Dtun is less than 1.0

and volume loss is equal or more than 1% For pile with Lp/Htun < 1.0, Jacobsz et al (2002), Kaalberg et al (2005) and Selemetas et al (2005) have defined their own zone of influence as shown in Figures 2.12, 2.14 and 2.15 respectively The defined zones are similar Furthermore, Cheng et al (2004) showed in the 3-D FE model that pile head settlement is in a decreasing trend when Lp/Htun is increased (see Figure 2.18a)

2.4.3 Pile axial force

Loganathan (1999) in the centrifuge tests shows that pile axial force is maximum at tunnel springline level when Lp/Htun is greater than 1 and at pile base level when Lp/Htun is equal to or less than 1 This trend agrees with those reported in numerical analyses where the axial force is increasing with depth due to the downdrag caused by tunnelling For Lp/Htun greater than 1, the pile length below tunnel springline would be subjected to positive skin friction to support the downdrag (i.e negative skin friction) In terms of magnitude, the induced axial force is greatest for

Lp/Htun < 1 (Figure 2.18b) The maximum axial force for volume loss of 1% is 180kN This corresponds to only 7% of the ultimate pile capacity However, the trend is totally different when compared to studies by others For example, Mroueh & Shahrour (2002) showed no clear trend of variation with Lp/Htun whereas Cheng et al (2004) showed that the axial force is consistently increasing with increasing Lp/Htun (see also Figure 2.18b) Besides, it can be observed that the magnitude of axial force is significantly small in the centrifuge tests for the similar details of tunnel and pile Furthermore, a pile could be subjected to tensile force particularly near the pile head This is likely to be caused by the type of restraint at the pile cap

2.4.4 Pile lateral deflection

Centrifuge tests by Loganathan (1999) showed that maximum lateral deflection of a pile occurs at the pile base regardless of Lp/Htun This contradicts field observation by Lee et al (1994) where the maximum occurs at the tunnel springline level for pile with Lp/Htun greater than 1 In terms of magnitude, the lateral deflection is largest when Lp/Htun is equal to 1 (Figure 2.18c) Besides, the magnitude of greenfield soil movement at pile position is similar to pile lateral deflection A magnitude of 7.5mm was observed for volume loss of 1% However, Lee et al (1994) only observed a magnitude of 7mm for volume loss of 2% The difference could be due to many other different factors such as pile diameter, pile-tunnel distance and soil condition Cheng et al (2004) also observed a similar magnitude of lateral deflection

2.4.5 Pile bending moment

In Loganathan (1999), bending moment was observed to be the maximum near pile base for varying Lp/Htun Besides, the maximum bending moment occurred for Lp/Htun of 1 (Figure 2.18d) For volume loss of 1%, the maximum bending moment is observed to be 90kNm This corresponds to only 12% of the ultimate moment capacity (Mult) The author also showed that for volume loss of up to 10%, the bending moment only reached 75% of Mult Mroueh & Shahrour (2002) in their FE analyses shows that the maximum bending moment varies similarly as the centrifuge tests by Loganathan (1999) where the maximum occurred for Lp/Htun at approximately

1 Despite that, Cheng et al (2004) showed that the maximum bending moment increases with

Lp/Htun (see also Figure 2.18d) The authors also concluded that bending moment would be negligible when Xpile/Dtun is greater than 2

In actual 3-D tunnel advancement, pile would be subjected to bending in transverse as well as longitudinal direction Most of the previous studies have been focusing on transverse bending due

to the simulation of plane strain tunnel However, Mroueh & Shahrour (2002) have carried out

3-D tunnel advancement simulation in FE and showed that the transverse bending moment is approximately three times the longitudinal bending moment

2.4.6 Pile group effect

To-date, only a few studies have been carried out on effect on pile group due to tunnelling Loganathan (1999) simulated a 2x2 pile group in the centrifuge tests From comparison with a single pile of the same distance to tunnel, the front pile was subjected to a reduction of approximately 16% and 22% of bending moment for pile with Lp/Htun of 0.86 and 1.00 respectively However, the bending moment increased by 40% for pile with Lp/Htun of 1.20 This inconsistency was not observed in the boundary element analysis carried out by the same authors reported in Loganathan et al (2001) In the front pile of the pile group, reduction of 12%, 29%, 6% and 15% were observed in pile head settlement, axial force, lateral deflection and bending moment respectively when compared to a single pile of the same distance to tunnel Similarly to the rear pile in the same pile group, reduction of 10%, 43%, 0% and 21% were observed respectively To be noted, lateral deflection for the rear pile was subjected to no reduction

Besides, Mroueh & Shahrour (2002) also reported in their 3-D FE analyses that reduction of 20% and 3% were observed for maximum axial force and bending moment in the front pile of a 2x2 pile group For the rear pile, much higher reduction were observed; 60% and 45% respectively Contrary to the above, Kitiyodom et al (2004) in their numerical analyses showed that almost negligible reduction was noted when the pile slenderness ratio (Lp/Dpile) was large (i.e 25)

Furthermore, when the Lp/Dpile was reduced to 5, a mix of positive and negative group effect can

be obtained The different observations lead to the inability for engineers to rely on the positive group effect as described in some of the reported studies

2.4.7 Multiple-tunnel advancement effect

Most of the tunnels particularly for urban railway transportation are usually constructed in a pair

or more to cater for the different bounding lines So far, there is no reported data looking into the effect of multiple-tunnel advancement on nearby pile foundation

A review of the published literature on the effects of tunnel construction on nearby pile foundation has been reported in this chapter A series of case histories, laboratory tests, field studies together with observations from numerical analyses were discussed Most of the case histories reported do not have in-pile instrumentation, and therefore, does not allow further investigation on pile response In all the laboratory tests carried out, plane strain tunnel was simulated and hence no 3-

D tunnelling effect to be studied In numerical analysis, the 2-stage approach was found to be less time consuming but does not adequately simulate pile-soil interaction 3-D FE analysis is advantageous over 2-stage approach owing to its capability to simulate 3-D tunnel advancement and pile-soil interaction However, the simulation is time consuming In 2-D FE analysis, the accuracy is subjected to the assumption of modification factor applied to the pile Finally, with the limited studies that have been carried out to-date, a general agreement among all the investigations could not yet be obtained Furthermore, contradictory observations were found in the study

CHAPTER 3

CASE STUDY: TUNNELLING ADJACENT TO PILE FOUNDATION FOR THE CONTRACT C704 - GROUND CONDITION AND FIELD MONITORING

3.1 Introduction

This chapter presents a unique case history in Singapore on the monitoring of the effects of tunnel construction adjacent to full-scale working piles As part of the construction of the Mass Rapid Transit (MRT) North East Line (NEL) Contract C704, a piled-viaduct bridge was planned in conjunction with the shield tunnel advancement Due to the early planning, in-pile instrumentation was possible and some strain gauges were installed along the pile reinforcement cage to determine the axial force and bending moment developed during tunnelling The instrumentation was implemented as part of the design of the piled-viaduct bridge and the tunnels in the same contract

An overview of the project is first highlighted followed by descriptions of the geological and ground conditions, design and construction details Subsequently, the methodology adopted in interpreting the instrumentation data is discussed Then, the monitoring results are presented, followed by comparison with previous works reported by others Before the concluding remarks, the design charts proposed by Chen et al (1999) were used to compute the pile responses and compared with the measured results

3.2 Back ground and overview of the project

The MRT NEL is Asia’s first fully automated underground rail system (Krishnan, 2000) Being the third major MRT line in Singapore since the completion of the main MRT network in 1990, the 22km long track is fully underground and consisted of 16 stations (Figure 3.1)

Contract C704 of NEL involved the construction of two cut and cover stations i.e Woodleigh Station and Serangoon Station, 992m twin tunnels between these stations (Ser-Wdlh) and 1522m twin tunnels from Serangoon Station to Kovan Station (Ser-Kov) The tunnels were bored with two Earth Pressure Balance machines (EPBM) Besides the tunnels and stations, the contract also included the construction of a 1.9km long dual-lane viaduct bridge

The viaduct bridge consisted of 2 abutments and 39 piers and was constructed in parallel alignment with the new twin tunnels configuration along the road Figure 3.2 shows the relative position of the tunnels alignments and bridge viaducts The piers are supported by pile groups of four to six bored piles of 1.2m and 1.8m in diameter Along the alignment, the piles are located between the twin 6.5m diameter tunnels The distance was as close as 1.6m between the tunnel extrados and pile edge Basically, there were two options for the construction sequence:-

Option 1 - Tunnels to be constructed first followed by the piled bridge viaduct

Option 2 - Piled bridge viaduct to be constructed first followed by the tunnels

The key concern of Option 1 was the effects of bored pile installation on tunnel lining Little

research has been reported on these effects and this option was ruled out to avoid additional

loading on the lining Option 2 was chosen as the understanding on this interaction is better compared to Option 1 Studies have previously been carried out using centrifuge test and finite

element model (see Chapter 2) Besides, Mair (1993) reported a tunnel in London which was

successfully constructed as near as 1m spacing between pile and tunnel without adverse effect

3.3 Geology and ground conditions

As seen in Figure 3.3, the tunnels in C704 were driven through two different geological formations, namely the Bukit Timah Granite and Old Alluvium However, the description of the geology and ground conditions will be limited to the Bukit Timah Granite where the instrumented piles were located The classification of its weathering grade is based on the widely adopted system in Singapore by Dames and Moore (1983) In the classification (Table 3.1), the Granite material was categorised into four groups which was in contrast to the six grades of the British Code of Practice (BS5930, 1981) The groups; G1, G2 and G4 represent the equivalent Grade I &

II, Grade III & IV and Grade V & VI of BS5930 respectively Group G3 is an additional category

to describe bouldery soil of variable weathering The G4 material is predominantly reddish brown, sandy silty clay or clayey silt The G1/G2 interface materials are highly permeable and exhibit sandy behaviour (Knight-Hassell & Tan, 2000) Being one of the oldest formation in Singapore, the Bukit Timah Granite was formed during the lower to middle Triassic period (Leong et al., 2003) The primary weathering is due to chemical decomposition and followed by secondary weathering which is a laterisation process

Figures 3.4 and 3.5 show respectively the basic and strength properties of G4 material measured at C704 The bulk unit weight, γbulk ranges from 17kN/m3 to as high as 22kN/m3 and is typically increasing with depth The Atterberg limit test on the soil type shows that the water content was generally nearer to the plastic limit which indicated the soil as a stiff material Plasticity index ranges from 10% to 30% The fine grain (i.e clay and silt) as measured by particle size

distribution shows a wide range, which was attributed to the varying degree of weathering Generally, the residual soil can be further categorised into two sub groups, i.e Group 1 is coarse grained soil and Group 2 is fine grained soil (Poh et al., 1985) But the average of 60% fine grain indicated the material is predominantly in Group 2

Strength properties as measured in triaxial test show that the average angle of shearing resistance, φ’ is 30o and the apparent cohesion ranges from 0 to 80kPa As summarised by Leong et al (2003), various researchers (Dames & Moore, 1983; Poh et al., 1985; Yang & Tang, 1997; Tan et al., 1988; Rahardjo, 2000; KarWinn et al., 2001; Zhou, 2001) have found that for Bukit Timah Granite residual soil, φ’ could lies between 13o and 40o but with an average approximately 29o to

32o Similarly, the c’ was found to range from 0 to 50kPa with average value lying between 9 and 26kPa Despite the wide scatter of data, both the reported values and the data found here are similar and consistent The undrained shear strength, Cu is increasing with depth but generally range from 20 to 200kPa The range of values is consistent with data reported by Poh et al (1985) Standard penetration test carried out in boreholes shows an increasing SPT-N value with depth The SPT-N value of 100 was reached at a depth ranging from 15m.b.g.l to 60m.b.g.l The high SPT-N encountered at shallower depth is due to shallower bedrock at certain location Typically, the ground water table was located at 3m.b.g.l

3.4 Design and construction details

The pile foundations supporting the bridge viaducts were designed to derive the geotechnical capacity from the shaft below the tunnel springline and the base resistance Negative skin friction along the pile shaft above tunnel springline was not considered in the design.All piles were base grouted to avoid soft toe problem These piles were designed to terminate at soil layer with SPT-N

>100 or bedrock whichever is encountered first To take into account the effects of tunnelling into the design, 2-D FE analysis was carried out

The piles were installed to depths ranging from 28 to 69m.b.g.l and generally varied in length due

to the degree of weathering of soil properties (Please refer to Lim (2003) for more information on

soil profile) The study here will only focus on piles at Piers 11, 14, 20, 32, 37 and 38 where

instrumented data are available Piers 11, 14 and 20 were located between Ser-Wdlh section where

as Piers 32, 37 and 38 were between Ser-Kov section At Piers 11, 14 and 20, the G4 material was

encountered down to 60m below ground surface without encountering bedrock (see Figure 4.9)

At a depth of 60m, harder G4 material exists i.e SPT-N>100 The G4 material with SPT-N ranging from 15 to 50 was encountered in the tunnelling zone However, at Piers 32, 37 and 38, the G4 material encountered was much shallower and typically down to a depth of 30m only followed by G1 material The G2 and G3 material were not apparent and a direct transition from G4 to G1 was observed The soil at the tunnelling zone was much stiffer at Piers 37 and 38 with SPT-N>100 where as at Pier 32, SPT-N<15

Bored piles of 1.2m diameters were installed to support Piers 11, 14, 20 and 32 where as 1.8m diameters were used at Piers 37 and 38 All the piers are supported by a group of four piles except for Pier 32 with five piles The piles at Piers 11, 14 and 20 are approximately 60m long where as piles at Piers 32, 37 and 38 ranges from 30m to 36m The tunnels were constructed close to the piles with clear distance ranges from 1.6m to 4.4m The typical relative position of the piles and the tunnels at Pier 20 is as shown in Figure 3.6 Reinforcement bars were used along the pile length and extend down to 5m below the tunnel invert The typical reinforcement of up to 20T25 longitudinal bars and T16 links at 175mm centre to centre were provided and corresponds to 0.87% area of pile Concrete of Grade 45 was used to cast the piles A pile cap of 1.5m thick was

located at a depth of approximately 1.5m below ground surface The typical dimension of pile cap was 5.3m x 5.3m for the 1.2m diameter piles and 7.7m x 7.7m for the 1.8m diameter piles Table 3.2 summarised the details of the instrumented piles

For the tunnelling work, EPBM was chosen owing to the highly variable ground conditions Two EPBMs, i.e EPBM1 and EPBM2 were used to bore the SB and NB tunnels respectively Both the EPBMs were launched from Serangoon Station and were advanced towards Woodleigh Station The tunnelling work started in April 1999 Upon breakthrough at Woodleigh Station, the EPBMs were transferred back to Serangoon Station and continued advance towards Kovan Station The

SB tunnel was first driven and followed by the NB tunnel which was approximately 200 to 300m behind the SB tunnel In sequence, Pier 20 was first reached and followed by Pier 14 and Pier 11

in the Ser-Wdlh section In the Ser-Kov section, Pier 32 was first reached and followed by Pier 37 and 38 In general, the SB and NB tunnels were located in almost the same level with the depth ranging from 16 to 29m.b.g.l

Knight-Hassell & Tan (2000)reported on the actual encounters during the tunnel drive in C704 In the first 600m of the drive for Ser-Wdlh section, G4 material was consistently encountered, after which G1/G2 interface material was encountered shortly before G1 material The G1 material was encountered for 40m before the G1/G2 material was encountered again Subsequently, full face G4 material was encountered again and followed by Old Alluvium (OA) and continued all the way for 350m until it reached Woodleigh Station At Ser-Kov section, soil condition similar to the tunnel drive for Ser-Wdlh section was encountered The G4 material was encountered initially followed by G1/G2 material and G4 material again G1 material was not encountered during the tunnel drive Beyond approximately 500m of the drive, OA material was consistently encountered all the way to Kovan Station An average face pressure of 150kPa was maintained during the

drive However, the soil is likely to have a substantial stand-up time even without face pressure (Shirlaw et al., 2003) The average progress rate was 4m/day and 9m/day respectively for Ser-Wdlh and Ser-Kov sections Furthermore, progress rate of up to 21m/day was achieved in some of the sections The differences were largely due to the change of soil weathering and breakdown in EPBM Tail void grouting was maintained at an average of 4.5m3 with the injecting pressure of 250kPa

3.5 Construction sequence

In C704, construction activities have been coordinated such that the bridge viaduct construction does not interfere with the tunnel advancement or in another word, affect the tunnel lining Therefore, all piles have to be installed prior to the EPBM advancement Piling work started in late 1998 At the time of the first EPBM reaching the piers, approximately 7.5 months, 5 months,

3 months, 11 months, 3 months 2.5 months has elapsed at Piers 11, 14, 20, 32, 37 and 38 respectively since piling Consolidation due to piling was presumed to have completed prior to EPBM advancement owing to the high permeability of the residual soil (i.e 1x10-7 m/s)

Besides, various levels of casting for the viaduct bridge were in progress at different piers during EPBM advancement At Pier 11, the construction had already reached flare head when both the EPBMs were adjacent to the pier At Pier 14, the pile cap had not been cast during the advancement of EPBM1 However, during the EPBM2 advancement, the entire stem pour has been completed At Piers 20, 32, 37 and 38, construction had only reached pilecap level prior to the EPBMs advancement Table 3.3 summarises all the construction stages

3.6 Instrumentation programme

Extensive instrumentation was installed along the tunnel route to monitor the surrounding ground during tunnel advancement and to verify the design Over 800 settlement and building markers, 48 inclinometers, 12 deep level extensometers, 10 tiltmeters, 26 tape extensometers and 55 piezometers/standpipes were installed along the tunnel route (Knight-Hassell & Tan, 2000) In this chapter, only selected instrumentation data are studied and presented

One of the important instrumentation was the in-pile strain gauges to monitor the pile response during tunnel advancement Piles at six piers i.e Piers 11, 14, 20, 32, 37 and 38 were instrumented with vibrating wire strain gauges cast into it In each pier, two piles denoted as ‘front’ pile (i.e nearest to the SB tunnel) and ‘rear’ pile (i.e nearest to the NB tunnel) were instrumented (see Figure 3.6) The strain gauges were located at four different levels spaced at 5m apart which correspond approximately to the tunnel springline, invert, crown and 5m above the crown Strain gauges were installed at 4 sides (i.e 2 pairs) in each level with each pair located parallel and perpendicular to the tunnel The strain gauges were spot welded on 10mm diameter sister bars before being covered by rust protectant and epoxy The strain gauges were used to calculate both the bending moment (i.e transversely and longitudinally) and axial force Besides, instrumentation such as inclinometer, magnetic extensometer, piezometer and settlement marker were installed around the piles Figure 3.6 shows the typical instrumentation layout at Pier 20 Similar instrumentation arrangement was also implemented at other piers

3.6.2 Interpretation of data

3.6.2.1 Assessment of surface settlement

One of the important parameters to assess the tunnel performance is the volume loss The measured surface settlement was converted to volume loss using Gaussian normal distribution as proposed by Peck (1969) Accordingly, Equation (3.1) is used

tun L

A

S i

V 250 max

where Smax is the maximum surface settlement on the tunnel axis (m) and Atun is the sectional area of tunnel (m2) The point of inflexion, i is defined by equation (3.2) where k is the

cross-trough width parameter which depends on the soil type and Htun is the tunnel depth (m.b.g.l.) It is

recognised that k of 0.5 would be most suitable representative of clay (Mair et al., 1993) and will

be used here The variation of surface settlement (s) with distance (x) away from tunnel axis can

3.6.2.2 Assessment of instrumented pile

To obtain the axial force (N), Equation (3.4) is adopted taking into account the average strain of the four strain gauges at each level

( )

4

2 1 2

x pile pileA E

where Apile is the cross section area of pile (m2), Epile is the Young’s modulus of pile (kN/m2) and ε

is the measured strain (+ve = tension and -ve = compression)

It should be noted that the piles are subjected to both vertical and horizontal loadings due to the three dimensional nature of tunnel advancement These produce non-uniform strain distribution in

a pile even at the same level For demonstration, a typical strain distribution on one of the

instrumented piles at Pier 20 is presented in Appendix A (see Figure A.2a) The averaging strain

(from Equation 3.4) is compared with strain from each individual strain gauge at the same level The large variation of strain particularly at the depth near tunnel springline is a consequence of the bending effect

The bending moment of a pile can be divided into two directions, namely longitudinal and transverse (see Figure 3.6) The pair of strain gauges perpendicular to the tunnel advancement is used to calculate the transverse bending moment where as the other pair in the parallel direction represents the longitudinal bending moment The two pairs of strain gauges in each direction respectively are interpreted from Equations (3.5) and (3.6) It should be noted that the bending moment interpreted is unfactored

y I

E

pile pile

One of the most important parameters in interpreting the axial force and bending moment of the pile is its Young’s modulus Selection of an appropriate value is not straight-forward since the concrete is known to be non-linear material and strain dependent Moreover, concrete tends to undergo creep, shrinkage and relaxation or in another word, time dependent In the geotechnical profession, these variations are usually ignored when interpreting or analysing concrete pile The pile is treated as a linear elastic material In order to take into account the effect, some methods such as tangent modulus method (Fellenius, 1989) and approximate method (ACI, 1989) to derive the Young’s modulus were investigated The results are presented in Appendix A The approximate method (ACI, 1989) yielded a comparable result with the lower bound range of the tangent modulus method and therefore, verified the accuracy of the method Ultimately, the approximate method is adopted for all the subsequent interpretation for simplicity Besides, moment of inertia (Ipile) is another variable that could affect the interpretation of bending moment

A preliminary check on the appropriate value was performed as presented in Appendix B It can

be observed that the bending moment in all piles mainly stayed within the cracked moment envelope and therefore, justified the applicability of gross moment of inertia in the interpretation

of bending moment

3.7 Monitoring results

3.7.1 Ground surface settlement

In a detailed investigation of the three-dimensional effects, it is important to identify the soil movement due to different phases of the tunnelling process In general, the settlement due to shield tunnelling can be derived from the face loss during tunnel advancement, shield loss due to

over-cut, tail void closure and consolidation (Shirlaw et al., 2003) Although pressure is applied at the tunnel face, it is inevitable that face loss might still occur It is therefore, necessary to maintain adequate pressure so as not to cause settlement or heaving of the ground Beyond the tunnel face, over-cut may result during the advancement of the shield machine The jacking forward will cause pitching or yawing of the machine and create a larger hole than the shield machine Subsequently, after the lining installation and tail void grouting, the shield machine advancement would expose the lining and grout to the soil Depending on the quality and how rapidly the grout could develop its strength to support the surrounding soil, further ground loss can occur The three types of losses

as described above are termed the immediate settlement as it happens within a short period of time Thereafter, consolidation will take place attributed to the generation of excess pore pressure during the tunnelling

Figure 3.7 shows the maximum surface settlement as measured directly above tunnel axis during the advancement from Serangoon Station to Woodleigh Station As can be observed, the settlement varies along the tunnel axis even though the same soil medium (G4 material) was encountered The ratio of the maximum to minimum measured is approximately 4.4 The variation

is probably due to the varying face pressure, degree of over-cutting and also the effectiveness of the tail void grouting Furthermore, at approximately 600m drive from Serangoon Station, the weathered interface material of G1/G2 was encountered which caused major ground losses Most

of these losses were grouted prior to migrating towards the ground surface (Shirlaw et al., 2003) and therefore, not reflected by the surface settlement markers

The surface settlement measured at Pier 20 during SB tunnel advancement is shown in Figure 3.8a Gaussian curve as described in Section 3.6.2.1 is fitted to the measured data and good

agreement was obtained The trough width parameter (k) of 0.5 assumed was found to be

adequately representing the soil Immediate settlement as measured corresponds to volume loss of

1.38% Further comparison with the tunnel advancement record allows the contribution from each phase of the tunnelling process to be obtained This was done by comparing the measured settlement trough with the position of EPBM Approximately 49% out of the total volume loss was due to face loss, followed by 28% of shield loss and 23% of tail void closure Face loss is a major contribution towards the surface settlement This is consistent with the observation reported

by Shirlaw et al (2003) where face loss was found to be the largest contribution to volume loss in Singapore tunnelling

Subsequent development of the surface settlement profile due to the second tunnel (NB tunnel) constructed side by side was also measured and presented in Figure 3.9 The figure shows the measurements of two points directly above SB and NB tunnels For convenience, the date is converted to ‘Day’ where Day 0 refers to 1 April 1999 (i.e the starting day of tunnelling work for C704) as the benchmark reference Immediately after the SB tunnel passed Pier 20, up to 5mm settlement was measured at the point directly above NB tunnel After about 50 days, the NB tunnel passed the same section and induced an additional 21mm settlement The point above SB tunnel settled another 9mm due to NB tunnel, which is much higher than that observed (i.e 5mm) above NB tunnel during SB tunnel advancement However, the final settlement at the two points was very similar Generally, the volume loss in each tunnel was slightly different A deduction of the volume loss merely due to NB tunnel was obtained by fitting Gaussian curve to the settlement trough by initially resetting the settlement due to SB tunnel to zero The result due to NB tunnel is shown in Figure 3.8b A total volume loss of 1.67% is calculated with the contribution of face loss, shield loss and tail void closure to be approximately 26%, 33% and 41% respectively Contrary to SB tunnel, the main contribution of surface settlement in NB tunnel was from the tail void loss Besides, the higher volume loss in NB tunnel compared to SB tunnel could probably be due to the variation of face pressure or inconsistency in tail void grouting In spite of that, the variation could also be due to SB tunnel which is constructed near to each other at approximately

2.5Dtun from centre to centre and caused disturbance to the soil, i.e reduction of soil stiffness Chapman et al (2004) and Ou et al (1998) also observed a higher settlement for the second tunnel constructed side by side

The settlement due to SB and NB tunnels was superimposed to allow the final settlement to be estimated as shown in Figure 3.10 The figure also revealed the effect of pile cap on the soil surface settlement In between the two tunnels, two settlement points were located on soil above the pile cap (see Figure 3.6) which restrained the soil movement Approximately 5mm settlement was apparent and could indirectly represent the settlement of pile cap

Similar to Pier 20 as discussed above, the volume loss is computed from the surface settlement markers at other instrumented piers location Table 3.4 summarises the volume loss computed at Piers 11, 14, 20, 32, 37 and 38 due to both the SB and NB tunnels The volume loss ranges from

as low as 0.32% to as high as 1.45% As can be noted, the volume loss is either similar or higher for NB tunnel compared to SB tunnel except at Pier 38

3.7.2 Subsurface soil movement

3.7.2.1 Vertical soil movement

The monitoring results from magnetic extensometers installed near the pile foundations as shown

in Figure 3.6 are discussed here Figure 3.11 shows the two readings at Pier 20 As the tunnel was progressively driven near the monitoring points, fluctuation of the soil vertical movement was observed This could be due to the influence of the shield tunnelling process such as face pressure, shield advancement and tail void grouting pressure